Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

Chapter 8 - Building Types S1/S1A: Steel Moment Frames 

8.1 Description of the Model Building Type 
Building Type S1 consists of an essentially complete frame assembly of steel beams and 
columns. Lateral forces are resisted by moment frames that develop stiffness through rigid 
connections of the beam and column created by angles, plates, and bolts, and/or by welding. 
Moment frames may be developed on all framing lines or only in selected bays. It is significant 
that no structural walls or steel braces are provided. Floors are cast-in-place concrete slabs or 
concrete fill over metal deck. These buildings are used for a wide variety of occupancies such as 
offices, hospitals, laboratories, and academic and government buildings. Figure 8.1-1 shows an 
example of this building type. 

Building Type S1A is similar but has floors and roofs that act as flexible diaphragms such as 
wood or untopped metal deck. One family of these buildings is older warehouse or industrial 
buildings, while another more recent use is for small office or commercial buildings in which the 
fire rating of concrete floors is not needed. 

Figure 8.1-1: Building Type S1: Steel Moment Frames 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

Variations Within the Building Type 

The use of structural steel in building construction started in the last decades of the 19th century. 
Lateral load resistance was initially provided by masonry infill. Additional resistance may have 
been provided by the encasement of the steel members in concrete, though these encased 
members were not designed with composite considerations. In the 1920s, the use of riveted 
connections introduced steel moment frames. Beam flanges and webs were joined to the 
columns by structural shapes, most commonly T-sections. Rivets had low strength and ductility, 
which limited the overall capacity of the frames. In the 1960s, high strength bolts replaced the 
rivets. Bolts were faster to install and permitted larger clamping forces, increasing the rigidity of 
the frames. Connections became smaller when cover plates bolted to the beam flanges and 
welded to the columns replaced T-sections in the 1950s. By the 1960s, beam flanges and webs 
were welded directly to the columns to create fully restrained connections, initiating the welded 
steel moment frame (WSMF) construction era. Shear tabs bolted to the beam webs and welded 
to the columns later replaced welded beam webs. These welded-flange and bolted-web 
connections were used extensively in the 1970s and 1980s and are now known as pre-Northridge 
connections. After the 1994 Northridge earthquake, it was concluded that these connections did 
not provide enough plastic rotational capacity for most seismic applications (FEMA, 2000c). 
Several types of connections have been developed and tested to address the flaws in the pre-
Northridge connections. The newer connections are designed to develop the full moment 
capacities of the beams and provide large inelastic rotation capacity. 

Floor and Roof Diaphragms 

Diaphragms associated with this building type could be either rigid or flexible. The typical rigid 
diaphragm found in modern buildings consists of structural concrete fill on metal deck. 
Diaphragm forces transfer to the frames through shear studs welded to the beams. Older steel 
buildings that were constructed before metal decks were commonly used may have concrete 
slabs or masonry arches that span between the beams. Flexible diaphragms include bare metal 
deck or metal deck with nonstructural fill. These are frequently used on roofs that support light 
gravity loads. Decks could be connected to the steel members with shear studs, puddle welds, 
screws, or shot pins. The steel members also act as chords and collectors for the diaphragm. 

Foundations 

There is no typical foundation for this building type. Foundations can be of any type including 
spread footings, mat footings, and piles, depending on the characteristics of the building, the 
lateral forces, and the site soil. Spread footings are used when lateral forces are not very high 
and a firm soil exists. For larger forces and/or poor soil conditions, a mat footing below the 
entire structure is commonly used. Pile foundations are used when lateral forces are extremely 
large or poor soil is encountered. The piles can be either driven or cast-in-place. Vertical forces 
are distributed to the underlying soil through a combination of skin friction between the pile and 
soil and/or direct bearing at the end of the pile; lateral forces are resisted primarily through 
passive pressure on the vertical surfaces of the pile cap and piles. 

8.2 Seismic Response Characteristics 
Steel moment frame buildings are generally flexible, but subject to large interstory drifts. The 
ductility of these buildings is achieved through yielding and plastic hinging of beams and/or 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

shear yielding of column panel zones at beam-column connections. This inelastic behavior 
allows moment frames to sustain many cycles of loading and load reversals. Historically, it was 
believed that large plastic rotations could be developed without significant strength degradation. 
Up until the Northridge earthquake, the performance of steel moment frame buildings was also 
believed to far exceed that of masonry and concrete buildings based on observations from 
previous earthquakes. 

The Northridge earthquake exposed severe deficiencies in WSMF connections. A significant 
number of the frames inspected after the earthquake exhibited visible cracking in the beam 
flange-to-column flange welds. In a few rare cases, the flanges completely fractured and the 
damage extended into either the shear tab or the column panel zone. Newer buildings, which 
relied on deeper beams with thick flanges and less redundancy, were discovered to be even more 
susceptible to this type of damage. In retrospect, a review of data from past earthquakes 
indicates that WSMF buildings rarely received close inspection following the event, and often 
these buildings were overlooked due to the more obvious damage in other types of structures 
(FEMA, 2000d). This oversight also applies to older steel buildings with riveted or bolted 
connections. These older buildings, though designed as moment frames, may have suffered 
limited damage due to the interaction of the steel frames with infill walls and concrete cores. 
Lastly, if steel damage was discovered after an earthquake, engineers often attributed this to poor 
construction quality (FEMA, 2000d). 

8.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
See Table 8.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 
Selected deficiencies are further discussed below by category. 

Global Strength 

The lack of global strength is caused by insufficient frame strength, resulting in excessive 
demands on the existing frames. Yielding or fracturing of the beams, columns, and/or 
connections could lead to excessive drifts. As a result, the building could be deemed irreparable 
after an eathquake. 

Global Stiffness 

Moment frames are much more flexible than other types of lateral force-resisting systems. Their 
flexibility could lead to excessive building drifts and interstory drifts. This is likely to cause 
structural damage to the connections and nonstructural damage to the partitions and cladding. 
Additional concerns include P- effects and pounding with adjacent buildings. Common 
rehabilitation measures include strengthening the existing frames or providing new vertical 
lateral force-resisting elements. 

Configuration 

Soft story conditions occur when stiffness from one floor to the next changes abruptly. This is 
common at ground floors of commercial and office buildings with tall first stories. It could also 
occur at mid-heights of five-story to fifteen-story story tall buildings that have not been designed 
for higher mode effects and near field motions. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

Table 8.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S1/S1A Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce Demand Remove Selected 
Components 
Global 
Strength 
Insufficient 
frame strength 
Moment frame 
Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Steel plate shear wall 
[8.4.8] 
Strengthen beams 
[8.4.3], columns [8.4.3], 
and/or connections 
[8.4.6], [8.4.9] 
Seismic isolation 
[24.3] 
Supplemental 
damping [24.4] 
Global 
Stiffness 
Excessive drift Moment frame 
Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Steel plate shear wall 
[8.4.8] 
Strengthen beams 
[8.4.3], columns [8.4.3], 
and/or connections 
[8.4.6], [8.4.9] 
Supplemental 
damping [24.4] 
Configuration Soft story Moment frame 
Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Steel plate shear wall 
[8.4.8] 
Re-entrant 
corner 
Moment frame 
Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Collector [8.4.4] 
Enhance detailing 
[8.4.3], [8.4.4] 
Load Path Missing 
collector 
Add collector [8.4.4] 
Inadequate 
shear, flexural, 
and uplift 
anchorage to 
foundation 
Embed column into a 
pedestal bonded to other 
existing foundation 
elements [8.4.5] 
Provide steel shear 
lugs or anchor bolts 
from base plate to 
foundation [8.4.5] 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

Table 8.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S1/S1A Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce Demand Remove Selected 
Components 
Load Path 
(continued) 
Inadequate out-
of-plane 
anchorage at 
walls connected 
to diaphragm 
Tension anchors 
[16.4.1] 
Component 
Detailing 
Inadequate 
capacity of 
beams, 
columns, and/or 
connections 
Enhance beam-column 
connection [8.4.6] 
Add cover plates or box 
members [8.4.3] 
Provide gusset plates or 
knee braces [9.4.1] 
Encase columns in 
concrete 
Inadequate 
capacity of 
panel zone 
Provide welded 
continuity plates [8.4.6] 
Provide welded stiffener 
or doubler plates [8.4.6] 
Inadequate 
capacity of 
horizontal steel 
bracing 
Provide additional 
secondary bracing 
[9.4.2] 
Strengthen bracing 
elements 
Reduce unbraced 
lengths 
Strengthen 
connections 
Diaphragms Inadequate 
in-plane 
strength and/or 
stiffness 
Collectors to distribute 
forces [8.4.3], [8.4.4] 
Moment frame 
Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Steel plate shear wall 
[8.4.8] 
Concrete topping slab 
overlay 
Wood structural panel 
overlay at flexible 
diaphragms [22.2.1] 
Strengthen chords 
[8.4.3], [8.4.4], and 
[22.2.2] 
Add nails at flexible 
diaphragms [22.2.1] 
Inadequate 
shear transfer to 
frames 
Provide additional 
shear studs, 
anchors, or welds 
[22.2.7] 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

Table 8.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S1/S1A Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce Demand Remove Selected 
Components 
Diaphragms 
(continued) 
Inadequate 
chord capacity 
Add steel members or 
reinforcement [8.4.3], 
[8.4.4] 
Excessive 
stresses at 
openings and 
irregularities 
Add reinforcement 
[8.4.3] 
Provide drags into 
surrounding diaphragm 
[8.4.4] 
Infill opening 
[22.2.4], 
[22.2.6] 
Foundations See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

Load Path 

Load path deficiencies in steel moment frame buildings include inadequate collectors and frame 
anchorage to foundations. In Type S1 buildings, seismic forces transfer from the diaphragm to 
the frame through shear studs welded to collectors or directly to the frame beams. The collectors 
or the connections to the frame may be too weak and insufficient to transfer these forces. 
Connections from columns to a base plate or pile cap have to resist shear, flexural and potentially 
uplift forces. Connections that cannot develop these frame forces do not allow the frame to 
develop its full capacity. 

Component Detailing 

The most common detailing deficiencies in steel moment frames are related to beam-column 
connections. Pre-Northridge welded moment connections consist of complete penetration flange 
welds and bolted or welded shear tabs. These connections were previously thought to be ductile 
but were found to fracture at small plastic rotations or even under elastic loads. Rehabilitation 
techniques primarily focus on forcing the yielding and plastic hinging to occur away from the 
joint to reduce stresses on the welds. This can be achieved with the use of reduced beam 
sections (RBS) or strengthening the section of a beam adjacent to the beam-column joint. These 
techniques are presented in Section 8.4.6 and discussed in detail in AISC Design Guide 12 (Gross 
et al., 1999) and FEMA 351 (FEMA, 2000b). Before welded moment connections became 
common, connections were either riveted or bolted. These types of connections are prone to net 
section fracture. 

Panel zones were previously thought to provide excellent ductility and strain hardening and were 
given increased predicted shear strength in the 1988 Uniform Building Code (FEMA, 2000c). 
The increased strength also meant larger inelastic demands in the panel zones. As a result, panel 
zones are subjected to yielding or buckling before adjacent members fully develop their 
capacities. Studies performed after the Northridge earthquake found that large panel zone 
deformations could increase the potential for connection failures (FEMA, 2000c). 

Welded column splices are vulnerable to fracture when subjected to large tensile loads (FEMA, 
2000b), since they generally use partial penetration groove welds and thus, are not designed for 
the full capacity of the smaller column. 

Diaphragm Deficiencies 

Common diaphragm deficiencies include insufficient in-plane shear strengths, inadequate 
chords, and excessive stresses at openings. Causes for these deficiencies could be due to lack of 
slab or fill thickness, lack of reinforcing steel in the slabs, insufficient connections to chord 
elements, and poor detailing at openings. See Chapter 22 for common rehabilitation techniques. 

Foundation Deficiencies 

Foundations that are inadequate do not develop the full capacity of the lateral force-resisting 
system. Their deficiencies result from insufficient strengths and sizes of footings, grade beams, 
pile caps, and piles. See Chapter 23 for common rehabilitation techniques. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

8.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
8.4.1 Add Steel Braced Frame (Connected to an Existing Steel Frame) 
Deficiency Addressed by Rehabilitation Technique 

Moment frame buildings that are insufficient to resist lateral forces or too flexible to control 
building drifts can be converted into braced frame buildings. 

Description of the Rehabilitation Technique 

The seismic performance of a building may be improved by adding braces to existing welded or 
riveted steel moment frames. Braces can be added without substantially increasing the mass of 
the building. Various concentrically braced frame (CBF) configurations should be considered, 
though some tend to perform much better than others in earthquakes. In addition, systems that 
meet the provisions for special concentrically braced frames (SCBF) are expected to exhibit 
stable and ductile behavior in large earthquakes. See Chapter 9 for a discussion of different CBF 
configurations and their behaviors. Moment frames are not commonly converted to eccentrically 
braced frames (EBF) due to complicated design and detailing issues that would be encountered. 
Different brace types can be used, including W-shapes, hollow structural sections (HSS), steel 
pipes, double angles, double channels, double HSS, and buckling-restrained braces. 

Design Considerations 

Adding braced frames to a moment frame building increases its stiffness considerably. The 
upgraded structure should be evaluated for higher lateral and overturning forces accordingly. 
The primary system performance issues are those associated with CBF systems, in which the 
most vulnerable elements are the braces and their end connections. Thorough knowledge of the 
existing material behaviors and strengths are necessary to ensure that new and existing elements 
to interact in the desired manner. Other design issues include the following: 

Research basis: No references directly addressing the addition of steel braced frames to moment 
frame buildings have been identified. 

Brace locations: Preference of brace locations should be given to existing moment frame bays to 
utilize the strengths of existing members, connections, and foundations. If this is not possible 
and braces are added at other locations, the existing moment frames should be considered when 
forces are distributed to the lateral force-resisting system. If the building drifts are large enough, 
deficiencies in the moment frames may still require mitigation whether the frames are included 
in the new lateral force-resisting system or not. 

Brace selection: Use compact and non-slender sections whenever possible to avoid premature 
fracturing or buckling of the braces during post-yield behavior. Issues related to conventional 
structural shapes are discussed in Chapter 9. Two particular brace types not common in older 
braced frame buildings are double HSS section and buckling-restrained braces. Double HSS 
sections can be used in configurations similar to double angles or channels (Lee and Goel, 1990). 
They provide reduced fit-up issues and smaller width-to-thickness ratios compared to a single 
HSS, resulting in increased energy dissipation capacity. The other type of brace, used in a 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

buckling-restrained braced frame (BRBF), is typically used in new buildings but has also been 
used successfully in new BRBF systems in existing buildings. One example of a brace used in a 
BRBF consists of a steel core inside a casing, which consists of a hollow structural section (HSS) 
infilled with concrete grout. Proprietary materials separate the steel core and concrete to prohibit 
bonding between the two materials. There are other buckling-restrained braces that do not use 
grout or additional separating agents between the steel and grout. The main advantage of these 
braces is the ability of the casing to restrain the buckling of the steel core without providing any 
additional axial force resistance beyond the capacity of the steel core. Provisions for new 
building BRBF design are included in the NEHRP Recommended Provisions (FEMA, 2003) and 
the AISC Seismic Provisions for Structural Steel Buildings (AISC, 2005b). 

Nonstructural issues: The addition of braces to an existing structure changes the architectural 
character of the building. Braces in exterior frames will be visible in buildings with clear glazing. 
At interior bays, braces have to be configured to avoid obstruction of existing corridors, 
doorways, and other building systems. Braces are also commonly exposed and incorporated into 
the interior architecture. For this particular option, note that gusset plates designed in accordance 
with the AISC Seismic Provisions for SCBF can be fairly large and should be discussed with the 
architect and tenants. If the braced frames are hidden in partition walls, the architect should be 
aware that these walls will be thicker than typical walls. Beams that are increased in size and 
new collectors may affect nonstructural components by reducing clear floor heights. These 
components typically include suspended ceilings, pipes, conduits, and ducts. Coordination with 
the architect and other trades should not be overlooked or underestimated. 

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 
available, and conducting comprehensive field surveys, the following issues should be noted: 

Connections: Gusset plates provide greater tolerances for the installation of braces than directly 
attaching the braces to the frame members. It may be impossible to completely eliminate 
welding at a braced frame connection in a seismic rehabilitation. Continuity and doubler plates 
on columns, stiffener plates on beams, brace-to-gusset plate connections, and most gusset plate-
to-frame member connections all require welding. A typical fully welded connection appropriate 
for use in SCBF is shown in Figure 8.4.1-1. If the brace capacity is governed by out-of-plane 
buckling, stable post-buckling behavior can be achieved by allowing the gusset plate to develop 
restraint-free plastic rotations. AISC recommends an offset of two times the plate thickness 
along the brace centerline, measured from the end of the brace to a line perpendicular to the 
nearest point on the gusset plate constrained from out-of-plane rotation. If gaps are provided 
between the gusset plate and concrete fill, then only the beam or column can act as constraints, 
as is the case in Figure 8.4.1-1. On the other hand, if concrete is placed directly against the 
gusset plate, then the slab can also act as a constraint. 

For low to moderate seismic applications, Figure 8.4.1-2 shows a more compact connection. An 
example of a W-shape used as a brace and welded directly to the beam and column is shown in 
Figure 8.4.1-3. The two latter connections do not allow for restraint-free plastic rotations out-ofplane 
and should be used primarily in situations where in-plane buckling of the braces govern or 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

Figure 8.4.1-1: HSS Brace at Existing Beam-Column Connection in SCBF 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

Figure 8.4.1-2: HSS Brace at Existing Beam-Column Connection in Ordinary CBF 

ductility demands are low. In addition, the connection shown in Figure 8.4.1-3 requires 
extensive field welding and could present fit-up issues if field dimensions are not verified on a 
case-by-case basis. 

Buckling-restrained braces are typically bolted to gusset plates. Their bolt configurations at 
these connections allow for very small tolerances. Double angle and double channel braces can 
be easily bolted to a gusset plate but space restrictions may limit the number of bolts and 
subsequently, the strength of the connection. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

Figure 8.4.1-3: HSS Braced at Existing Beam-Column Connection 

It may not always be practical or possible to locate a work point at the intersection of the 
centerlines of the beam and column, e.g. deep beams. This may deemed to be acceptable if the 
eccentricity is included in the connection design. 

Built-up brace members: While double angles, double channels, and double HSS offer 
advantages for installation, more stringent criteria apply to these members when used in SCBF. 
These include stitch spacing, member compactness, and strength of the stitches (AISC, 2005b). 

Reinforcing cover plates: HSS and pipe braces are subject to net section fracture at the gusset 
plate slots (Uriz and Mahin, 2004, and Yang and Mahin, 2005). This brittle failure mode can be 
eliminated by adding reinforcing cover plates to the sides of the HSS without the slots, as shown 
in Figure 8.4.1-1 and Figure 8.4.1-2. For pipes, the reinforcing plates can be oriented at right 
angles to the pipe and appear like stiffeners. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

Cost/Disruption 

This system could be cost-effective when compared to other alternatives for upgrading steel 
moment frame buildings. Costs will be less when existing moment frames are converted into 
braced frames to take advantage of the existing strength and stiffness of the frame members, 
connections, and foundations. Designs that are simple and details that are not overly 
complicated will also minimize costs. 

Costs can also be reduced if disruption is minimal during construction. Installing braces at the 
perimeter frames reduces logistical issues associated with working in confined spaces and 
temporary removal of the nonstructural elements. Noise associated with this type of work is loud 
and disturbing to the tenants if the building is occupied while the work is being performed. 

Construction Considerations 

The engineers involvement during the construction phase is critical during a seismic 
rehabilitation. The design of the retrofit scheme must not neglect the construction phase and 
should consider these issues at a minimum: 

Welding issues: A work environment in which the welder can perform quality welds is critical. 
This includes an environment with adequate space to properly operate welding equipment, 
adequate lighting, and a stable work platform for overhead welds, all while ensuring worker 
health and safety. Additional considerations include venting of welding fumes and fire 
protection. 

Removal of existing nonstructural elements: Exterior cladding and interior partitions must often 
be removed to deliver and install the braces to their final locations. Connection modifications at 
the roof level may warrant the removal of roofing and waterproofing. Installation of the 
connection to the underside of beams, column continuity and doubler plates, and beam stiffeners, 
will affect ceilings, lights, and other mechanical/electrical/pluming components. Almost all steel 
buildings will have some form of fireproofing around the members and connections. With older 
buildings, asbestos may be present in the fireproofing, which could require costly abatements to 
expose the members. Older buildings with items of historical significance may require 
additional coordination and effort to ensure that these items are removed and restored properly. 
Buildings constructed in the early 20th century commonly had steel members encased in 
concrete. 

Removal of existing structural elements: Slabs and metal decks must be chipped and cut away to 
install connections. Slabs oriented perpendicular to the beams require temporary shoring. 
Temporary openings in the slab not only require shoring but also consideration for how the 
openings will be closed. Beams are rarely removed but if unavoidable, the slabs supported by 
the beams need shoring and the columns supporting the beams may require bracing. 

Construction loads: Loads from construction activities vary and may be either temporary or 
permanent. Temporary loads could include the weight of construction equipment and patterned 
loading on perimeter framing when heavy cladding is temporarily removed. Permanent loads 
could be induced if connection stiffness is modified, such as converting a simple shear tab 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames 

support into a fully rigid moment connection, and when members are temporarily removed, 
causing forces to redistribute. 

Proprietary Concerns 

Braces used in BRBF are proprietary. There are a limited number of manufacturers of the braces 
used in BRBF. 

8.4.2 Add Concrete or Masonry Shear Wall (Connected to An Existing Steel 
Frame) 
Deficiency Addressed by Rehabilitation Technique 

Moment frames buildings that are insufficient to resist lateral forces or too flexible to control 
building drifts can be strengthened and stiffened by adding shear walls. The shear walls may be 
used alone as the new lateral force-resisting system or in conjunction with the moment frames. 

Description of the Rehabilitation Technique 

Shear walls can add considerable strength and stiffness to a structure. These walls could be 
considered as an alternative to adding braced frames if the existing beams and columns are 
incapable of resisting forces in the braced frame system. Concrete shear walls can be placed 
using conventional formwork or shotcrete can be used instead if skilled operators for placing 
shotcrete walls are available. Masonry shear walls, though typically weaker than their concrete 
counterparts, have the advantage of not requiring formwork when filling the cell cavities with 
grout. 

Design Considerations 

The addition of a shear wall to an existing steel frame forms a composite shear wall system. The 
horizontal shear forces are resisted by the wall elements and the vertical overturning forces are 
primarily resisted by the steel columns that become boundary elements. This system is almost 
always controlled by shear due to the substantial flexural strength provided by the steel column 
boundary elements. Though shear cracking and yielding of a wall is not as ductile as flexural 
hinging at the base of the wall, limiting the wall to small drifts prevents early loss of strength and 
stiffness degradation. Other design issues are as follows: 

Research basis: No references directly addressing the addition of shear walls to moment frame 
buildings have been identified. 

Design forces: The forces on the structure could increase significantly when this mitigation 
technique is employed due to the increased stiffness and mass of the walls. The entire structure 
should be reanalyzed, including all components that were previously determined to have 
sufficient capacity to resist the forces on the more flexible structure. In-plane forces due to the 
self-weight of the walls remain in the walls and do not increase demands on other elements. The 
out-of-plane forces, on the other hand, must be transferred through the diaphragms and collectors 
to other walls or frames. The design of the wall foundations depends on the magnitude of the 
forces and the soil properties. If overturning forces are greater than the available counteracting 
building mass or the soil is poor, some type of pile foundation will be necessary; otherwise, strip 
footings may suffice. 

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Techniques for the Seismic Rehabilitation of 
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Wall locations: Preference of shear wall locations should be given to existing bays of moment 
frames to utilize the layout of the existing collectors and the strengths of the existing members 
and connections. If this is not possible and shear walls are added at other locations, the existing 
moment frames should be considered when forces are distributed to the lateral force-resisting 
system. Deficiencies in the moment frames may still require mitigation whether the frames are 
included in the lateral system or not. 

Continuous walls vs. frames with infill walls: A continuous wall that encases the existing frame 
beams and columns, shown in Figures 8.4.2-1 and 8.4.2-2, respectively, provides the most robust 
and simple design. It may be desired to use a system that utilizes independent wall panels that 
appear similar to older style steel frame buildings with infill walls. The detailing and behavior 
would be much different from the older buildings though. The wall panels in each bay do not 
encase the beams and columns, as shown in Figures 8.4.2-3 and 8.4.2-4, respectively. One 
advantage of this system is that it may allow the elimination of shoring by not having to remove 
the metal deck though some slab concrete removal may still be required. Another advantage is 
that beams that have existing penetrations and the elements penetrating the beams would not be 
affected. The main drawback is that this system may be limited to use in buildings with lower 
seismic forces. The wall strengths would be limited by the number of studs that can be installed 
and the capacity of the beam webs. Another option is to encase the beams but not the columns. 

Coupled walls: Beams in existing bays of moment frames that are located between new walls 
will behave like coupling beams, particularly if the bay is relatively short. These beams, subject 
to high shears and moments, require ductile detailing. The level of coupling between the walls is 
a function of the stiffness of the beams, which can be determined by a computer model of the 
system. 

Nonstructural issues: The addition of walls to an existing structure changes the architectural 
character of the building. Walls at the exterior will be visible in buildings with clear glazing. At 
interior bays, walls have to be configured to avoid obstruction of existing corridors, doorways, 
and other building systems. Walls can be exposed and incorporated into the interior architecture 
or hidden in partition walls. The architect should be aware that these walls will be thicker than 
typical partition walls. Encased beams and new collectors affect nonstructural components by 
reducing clear floor heights. These components typically include suspended ceilings, pipes, 
conduits, and ducts. Encased columns reduce usable floor space. Coordination with the 
architect and other trades should not be overlooked or underestimated. 

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 
available, and conducting comprehensive field surveys, the following issues should be noted: 

Connection to existing frame: Positive connection should be provided to the existing frame to 
transfer seismic loads to the walls and to provide overturning resistance. Welded shear studs or 
reinforcing bars are typically used. In much older buildings where cast iron members still exist, 
it is difficult to weld to cast iron due to its high carbon content. Instead, holes for the reinforcing 
steel bars may have to be drilled in the members to rely on direct bearing for the load paths. 

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Figure 8.4.2-1: Cast-in-Place Concrete Wall at Existing Beam 

Wall reinforcing: Reinforcing steel in the walls should meet all ACI 318 (ACI, 2005) 
requirements, including the seismic provisions. Columns with positive connections to the walls 
could be counted as boundary reinforcing steel limited by the strength of the shear connections. 
Congestion issues may be encountered at the ends of the walls due to the presence of the 
columns, confinement steel, and anchorage for horizontal wall reinforcing steel. It may be 
necessary to drill holes through the columns for the horizontal steel. 

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Figure 8.4.2-2: Cast-in-Place Concrete Wall Encasing Existing Column 

Mechanical couplers: Consider using reinforcing steel couplers at areas of congestion. Note that 
diameters of the couplers could be as much as twice that of the bars and must meet ACI cover 
requirements for reinforcing steel. 

Effective wall thickness: At beam flanges, the wall thickness could be reduced considerably, 
which compromises the shear strength of the wall. There are several ways to add the shear 
strength of the wall at these locations. Shear studs can be added to the beams to transfer some 
forces through the beam web. Thickness of the wall can be increased at the beam. Extra shear 
reinforcing steel can be placed in the wall provided that ACI limits for reinforcing steel are not 
exceeded. 

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Figure 8.4.2-3: Discontinuous Wall at Existing Beam 

Wall connections at slabs: To avoid damaging the slab reinforcing steel critical for transferring 
diaphragm forces to the shear walls, the concrete slab could be chipped away without damaging 
the slab steel. This still allows for a monolithic wall construction with a construction joint most 
likely located at the top of slab. If the shear forces in the walls are sufficiently low, the entire 
slab may be preserved and merely roughened at the construction joint. Holes can be drilled in 
the slab to allow vertical reinforcing steel to pass through. 

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Figure 8.4.2-4: Wall at Existing Column 

Frame with infill walls: Since the purpose of this technique is to create a shear wall system, the 
detailing should meet this goal by providing continuous load paths for both the laterally induced 
shear and overturning forces. Welded shear studs or reinforcing bars should be provided along 
all wall panel edges to achieve this. The horizontal shear force transfers from one wall to 
another through the studs on the top and bottom flanges and the webs of the beams at each floor. 
Additional plates could be welded between flanges to reduce stresses in beam webs. The 
columns become the boundary elements of the shear walls. Their effectiveness in providing 
overturning resistance may be limited by the number of shear studs or reinforcing bars that can 
be installed. Consider welding one stud on top of another to increase the embedment depth to 
avoid pullout failure. 

Offset walls: Walls do not necessarily have to be centered on the existing beams and columns. 
By offsetting a wall and using the beam and column webs as the edge of the wall, shoring and 
metal deck removal only has to be performed on one side of the frame. An offset wall also lends 
itself to shotcrete construction, while walls with centered steel members have to be cast-in-place 
since it is not possible to place shotcrete on both sides of a steel member. 

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Cost/Disruption 

Adding concrete walls to an existing building is costly, though some savings can be found in the 
wall construction. Shotcrete walls are typically cheaper and faster to construct than conventional 
concrete walls due to the savings in materials and labor associated with formwork. Cost savings 
can be even greater if shotcrete is applied against an existing wall at a stair or elevator and 
mechanical shafts. CMU walls are generally less expensive than either shotcrete or cast-in-place 
concrete. New foundations are almost always required for new walls and could be extremely 
costly if deep foundations, such as drilled piers, are added. 

Installing new walls is disruptive to the occupants because of the noise and vibrations associated 
with construction. Even if tenants are relocated to parts of the building where the work is not 
being performed, vibrations associated with cutting, chipping, and drilling of concrete can 
transmit through the structure. The disruption can be reduced somewhat if the walls are installed 
at the perimeter. 

Construction Considerations 

The engineers involvement during the construction phase is critical during a seismic 
rehabilitation. The design of the retrofit scheme must not neglect the construction phase and 
should consider the issues below at a minimum. 

Shotcrete walls: The quality of a shotcrete wall is highly dependent on the skill of the nozzle 
operator. Building codes typically require preconstruction test panels constructed and reinforced 
similarly to the actual walls. This allows for inspection of the finished product, sawcutting to 
verify the quality of the shotcrete, and coring to determine strength. Additional test panels for 
overhead joints should also be requested to allow for cores to be taken to inspect the surface 
preparation and the joint bond. A series of ACI 506 publications provide general information, 
specifications, certification of nozzle operators, and evaluation of shotcrete (ACI, 1991, 1994, 
1995a, 1995b, 1998). 

Congestion: Congestion issues may be encountered at the ends of the walls due to the presence 
of the columns, confinement steel, and anchorage for horizontal wall reinforcing steel. Consider 
using mechanical couplers in heavily reinforced boundary elements and walls where lap splices 
are impractical. 

Concrete/shotcrete placement: The type of construction should be determined during the design 
phase because details for cast-in-place concrete and shotcrete construction are different. It may 
not always be possible to confidently use either type of construction; and, thus, the details 
creating such situations should be avoided. During construction, some locations require 
particular attention when shotcrete and concrete are being placed to ensure that all gaps are filled 

 tops of wall panels, k-region of steel sections, and construction joints. Figure 8.4.2-5 shows a 
pilaster at a steel column where shotcrete shadowing restrictions require the use of cast-in-place 
construction for the pilaster but still allow shotcrete construction away from the pilaster. See 
Section 21.4.5 for additional discussion concerning concrete and shotcrete construction. 
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Removal of existing nonstructural elements: Exterior cladding and interior partitions may have to 
be removed to deliver concrete formwork and other equipment. Walls that encase beams at the 
roof require temporary removal of the roofing and waterproofing. Installation of the walls will 
affect ceilings, lights, and other mechanical/electrical/pluming components. Nonstructural 
elements located within the frame bays being strengthened have to be moved for construction of 
the walls. Permanent relocation may be desired for some of these elements to minimize 
openings in the new walls. Older buildings with items of historical significance require 
additional coordination and effort so that these items are not damaged when temporarily 
removed and are restored properly. 

Removal of existing structural elements: See general discussion in Section 8.4.1. 

Construction loads: See general discussion in Section 8.4.1. 

Proprietary Concerns 

Mechanical couplers for reinforcing steel are proprietary. 

Figure 8.4.2-5: Combined Shotcrete and Cast-In-Place Construction 

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8.4.3 Add Steel Cover Plates or Box Existing Steel Member 
Deficiency Addressed by Rehabilitation Technique 

Frame members that are inadequate to resist the seismic demands are strengthened with cover 
plates or by adding side plates to W-shapes to create box sections. This reduces axial and 
flexural stresses in beams and columns and could also be used to increase the shear strengths of 
these members. 

Description of the Rehabilitation Technique 

Cover plates are welded to the outside of existing flanges, which in effect, increase the flange 
areas. When strengthening a beam, a cover plate is more commonly attached only to the 
underside of the beam since the presence of an existing diaphragm makes it difficult and costly 
to weld to the top of the beam or attach side plates to box the beam. However, for strengthening 
non-composite beams, top cover plates are virtually indispensable. Conversely for columns, 
plates are typically welded to the sides of W-shapes because it is much more effective to increase 
the axial and flexural capacities by converting a column to a box section and the shear capacity 
by essentially providing two additional webs. 

Design Considerations 

Thorough knowledge of the existing material behaviors and strengths are necessary for the new 
and existing elements to interact in the desired manner. If welding to the existing components, 
the carbon equivalent of these components need to be verified through as-built records or new 
testing to determine their weldability; see FEMA 356 (FEMA, 2000e) for further discussion of 
this issue. Other design issues include the following: 

Research basis: No references directly addressing the addition of steel cover plates to existing 
frame members have been identified. 

Beams: The flexural strength of a beam can be improved by welding cover plates to the bottom if 
there is a composite slab present. If there is only a bare metal deck, a cover plate on only one 
side of the beam may not be very effective. However, it could be useful for strengthening beams 
with large axial forces, primarily collector members. 

Columns: The overall capacity of a column is determined by axial-flexural interaction. Boxing a 
column decreases its slenderness and, therefore, increases its axial and flexural capacities. 
Whether the areas of the new plates can be directly included in computing these capacities 
depends on their continuity and detailing at a beam-column joint. Except for one side of the 
exterior columns, beams framing into the columns at each floor will disrupt the continuity of the 
new plates. 

Foundations: Where cover plates are added to the columns at their base, a reevaluation of the 
foundation system is warranted. It is not uncommon for frame columns to develop plastic hinges 
at their bases and thus, the increase in demand on the foundation may be greater than intended. 
The strength and stiffness of the base plate and the anchor rods should be evaluated and 
upgraded accordingly. A more in-depth discussion of column connections to the foundation is 

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provided in Section 8.4.5. Foundation upgrades are not common with this technique, but if 
required, refer to the chapter on foundations. 

Nonstructural issues: The beam and column upgrades hardly take up any additional space but 
minor modifications and temporary relocations may be required for some architectural and 
M/E/P elements. 

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 
available, and conducting comprehensive field surveys, the following issues should be noted: 

Beams: When a section of beam is strengthened, the cover plate typically attaches to the bottom 
flange with fillet welds, as shown in Figure 8.4.3-1. The length and size of the welds determine 
the shear flow between the plate and the existing beam, similar to interactions between other 
composite elements. 

Columns: Side plates used to create a box section can be attached to the column flanges with 
CJP, PP, or fillet welds, as shown in Figure 8.4.3-2. The fillet weld is the simplest to implement 
since it does not require additional preparations like the other welds, particularly the use of 
backing plates and beveling of the welded edges. The type of connection used may limit the 
shear flow between the plates and the existing column. 

Existing floors: Whether the columns are upgraded with cover plates or box sections, the new 
plates have to terminate at the existing beams and slabs. Though it is possible to achieve 
continuity by using additional plates or sections if necessary, this would create detailing and 
construction complexities and should be avoided. Furthermore, excessive welding around a 
beam-column joint could create undesirable residual stresses in the joint. 

Cost/Disruption 

Schemes that involve slab removal, work around a connection, and foundation work are costly. 
As typical with seismic upgrades, cost and disruption is minimized when schemes are kept 
simple. 

Construction Considerations 

See Section 8.4.1 for general discussion of welding issues, removal of existing nonstructural and 
structural elements, and construction loads. 

Proprietary Concerns 

There are no known proprietary concerns with this technique. 

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Figure 8.4.3-1: Cover Plate at Existing Beam 

Figure 8.4.3-2: Box Section at Existing Column 

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8.4.4 Provide Collector in a Concrete Fill on Metal Deck Diaphragm 
Deficiency Addressed by Rehabilitation Technique 

Beams that are inadequate to transfer collector loads are strengthened. 

Description of the Rehabilitation Technique 

Plates can be welded to various portions of the existing beam. Cover plates, discussed in Section 
8.4.3, are commonly attached to the underside of beams. To attach plates to top flanges, it may 
be simplest to orient the plates vertically and weld to the underside of the flanges, as shown in 
Figure 8.4.4-1, or similarly on top of the bottom flanges. Given all the possible options, the plate 
locations may be dictated by the eccentricities, continuity options at a beam-column joint, or the 
presence of nonstructural elements. 

Design Considerations 

Though the purpose of this technique is to strengthen a beam axially, it also changes the flexural 
properties of the beam by increasing its stiffness. It should be verified that this does not have 
unintended consequences, such as converting the beam into a frame member. Eccentricities in 
collectors are typically neglected. However, deep collectors and/or collectors with large forces 
could be subject to significant moments from these eccentricities. The eccentricities could be 
reduced if plates are attached to the top flanges but requires other detailing considerations at a 
beam-column joint. 

Detailing Considerations 

The new plates should attach to the columns with complete joint penetration (CJP) welds. 
Consider replacing the welds in the existing collector with high notch toughness welds in a high 
seismic region. Continuity plates directly aligned with the collector elements at a beam-column 
joint should be used as much as possible but may be offset if all eccentricities and their effects on 
the joint are considered. 

Cost/Disruption 

This technique is relatively inexpensive compared to other types of steel frame upgrades. The 
most expensive part of the structural upgrade is associated with the work at the beam-column 
connections. The nonstructural disruptions to the architectural and M/E/P elements are typical. 

Construction Considerations 

See Section 8.4.1 for general discussions of welding issues, removal of existing nonstructural 
and structural elements, and construction loads. 

Proprietary Concerns 

There are no known proprietary concerns with this technique. 

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Figure 8.4.4-1: Plate Collectors at Existing Beam 

8.4.5 Enhance Connection of Steel Column to Foundation 
Deficiency Addressed by Rehabilitation Technique 

Frame columns are subject to axial (including possible tension), flexural, and shear forces. To 
this end, columns with inadequate anchorage to the foundation limit the capacity of a frame. The 
columns could be part of an existing lateral force-resisting system that do not meet current 
standards or part of an upgraded system with larger forces resulting from increased stiffness. 

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Description of the Rehabilitation Technique 

Two methods are common for enhancing the column connection to the foundation. First, 
modifications at the base plate could include the addition of anchor rods, welding shear lugs to 
the base plate, and/or enlargement of the base plate. The other method is to encase the column in 
a concrete pedestal. It is also possible to use both of these methods together. The foundation 
system itself is not addressed here but is discussed in Chapter 23. 

Anchor rods can be used to resist tensile forces due to uplift and flexure. Holes drilled in the 
existing footing for these rods can be filled with a nonshrink high strength grout or chemical 
adhesive. Shear lugs may be used to transfer shear forces into the foundation. If the existing 
base plate is not large enough to accommodate the new rods or lugs, new plates can be welded to 
the existing plate to enlarge its area. This is also necessary if the allowable bearing stress is 
exceeded due to increased column compression. The increased base plate size leads to greater 
shear and flexural forces in the plate. It is likely that the plate will not be thick enough to resist 
these forces. Thus, stiffeners that act as supports can be welded to the base plate to reduce the 
plate forces. The stiffeners also provide additional load paths from the column to the base plate. 
Figure 8.4.5-1 shows some of these modifications. 

Base plate modifications may not always be practical or possible. Instead, the column can be 
encased in a concrete pedestal above the footing, as shown in Figure 8.4.5-2. Shear forces would 
transfer through direct bearing of the column against the pedestal. The pedestal could also be 
used to transfer uplift and flexural forces by relying on the existing base plate and other 
mechanisms, such as welded shear studs along the column. The pedestal itself should be detailed 
as a reinforced concrete column that meets all ACI requirements and seismic provisions. 

Design Considerations 

Thorough knowledge of the existing material behaviors and strengths are necessary for the new 
and existing elements to interact in the desired manner. A complete presentation of column base 
plate design can be found in AISC Design Guide 1 (Fisher and Kloiber, 2005). Key design 
issues include the following: 

Research basis: No references directly addressing the upgrade of connections of steel columns to 
foundations have been identified. 

Shear lugs: Columns that are in compression offer shear resistance through friction below the 
base plate. AISC recommends different friction coefficients depending on the location of the 
base plate with respect to the top of concrete. Additional shear resistance is required when the 
compressive force is not great enough or there is tension in the column. Though not common in 
retrofit applications, shear lugs, or simply plates welded to the bottom of a base plate, have the 
advantage of shallow embedment compared to anchor rods. The lugs also receive confinement 
from the base plate above and simplify the design by allowing anchor rods to only resist tension. 
The design of a shear lug is a function of the allowable bearing stress in the surrounding grout 
and flexural forces in the lug. 

Anchor rods: Tension in an anchor rod is developed through bond along its length and/or direct 
bearing through a hook, bolt head, or nut at the end of the rod. Though it is possible to drill a 

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hole large enough in an existing footing to accommodate a bolt head or nut, manufacturers of 
grouts and adhesives sometimes limit the size of the a hole for a given rod diameter in which its 
product can be used as an infill. In this case, the tensile force has to be developed entirely 
through bond. For bond development, the rod has to be threaded or some other means of 
mechanical anchorage has to be provided along the rod. Anchor rods are not typically used to 
transfer shear to the foundation because the mechanism for shear transfer is difficult to define 
and still subject to debate. If unavoidable, several factors should be considered, including 
bending of the rod through the oversized hole in the base plate and shear and tension interaction 
on the rod. If the column shear force is being distributed between both new and existing rods, 
note that the holes for the existing rods are likely to be oversized and welded plate washers 
should be verified or provided. 

Figure 8.4.5-1: Modified Base Plate to Increase Uplift Capacity 

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Figure 8.4.5-2: Concrete Pedestal at Existing Column 

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Stiffeners: The simplest way to reduce flexural stresses in a base plate is to add stiffeners that 
essentially provide multiple supports along the base plate. These stiffeners also assist in 
transferring uplift and flexural forces to the base plate and reduce the weld stresses at the base of 
a column. 

Nonstructural issues: Base plates are typically hidden in the slab and thus, the upgrade could 
remain hidden if it only involves adding anchor rods. Pedestals, on the other hand, require 
additional space and could present aesthetic issues. However, basements are often used for 
parking, storage, or as equipment locations, where aesthetic considerations may be negligible. 

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 
available, and conducting comprehensive field surveys, the following issues should be noted: 

Base plates: If the existing base plate can accommodate new anchor rods or lugs, burning holes 
in the plate is usually acceptable. Otherwise, enlargement of an existing base plate is 
accomplished by welding new plates to the existing plate. Only partial penetration (PP) welds 
can be used since the lack of backing precludes using complete joint penetration (CJP) welds 
unless concrete is removed to allow placement of backing plates. The depth of the PP weld 
limits the effective thickness of a plate. Also of note is that AISC relaxes its typical edge 
distance provisions if there is no lateral load on a base plate. The only recommendation is that 
enough edge distance remains such that the drill or punch does not drift when a hole is made. 
AISC Design Guide 1 suggests that one quarter of an inch is enough to meet this condition. 

Anchor rods: Detailing differs depending on if an anchor rod is designed to resist small shear 
forces and tension or tension only. For the latter, an oversized hole in the base plate and a thick 
plate washer over the hole is considered adequate. Since oversized holes are intended to allow 
for inaccuracies in anchor rod placement when concrete footings are placed, it should be possible 
to use smaller holes if the holes are being drilled. In addition to the design considerations 
mentioned above, anchor rods with shear forces require welding heavy plate washers to the base 
plate. The washers should have close-fit holes to minimize the movement required to engage all 
of the rods together. Nuts are sometimes welded to the washers but the mechanical properties of 
the nuts could lead to welding complications. 

Pedestals: Reinforcement of the pedestal is similar to a reinforced concrete column. Vertical 
bars provide tensile and flexural strength while hoops or ties provide confinement. At the top of 
footing, the shear is transferred through shear friction in the anchored reinforcing steel dowels. 
The concrete surface should be roughened, which also increases the shear friction values. If the 
pedestal is also used to transfer uplift and flexural forces, additional dowels are required for 
tension. A single dowel should not be considered to provide both shear and tensile resistance. 
One mechanism to transfer uplift and flexural forces from the column to the pedestal is the use of 
welded shear studs along the column. The size and spacing of the shear studs may determine the 
minimum pedestal diameter and height. 

Concrete anchorage: Installation of the anchor rods and dowels requires drilling into the existing 
footing. Consider using a scanning device to avoid damaging the reinforcing steel in the top 

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layer of the footing. Cores may have to be taken to confirm the concrete strength. Many 
commercial grouts and adhesives are available for bonding the anchor to the concrete. Verify 
that the selected product is appropriate for the specific anchor in a seismic application. On site 
inspection and testing of the anchors are mandatory since their performance rely heavily on the 
installation process. The grout and adhesive product vendors ICC Evaluation Service reports 
provide standardized installation procedures and anchorage capacities. 

Cost/Disruption 

The cost of a column connection to foundation upgrade is not very expensive. However, this 
technique is not usually performed only by itself. Costs would be small relative to an overall 
lateral force-resisting system upgrade and a foundation upgrade. Disruption could be minimal 
since typically, there are no tenants in the basement. Even if the basement were used for tenant 
access, such as parking, only a few columns would be affected at a time. 

Construction Considerations 

The engineers involvement during the construction phase is critical during a seismic 
rehabilitation. The design of the retrofit scheme must not neglect the construction phase and 
should consider these issues at a minimum: 

Welding issues: See general discussion in Section 8.4.1. 

Removal of existing nonstructural elements: This technique tends to be less disruptive of 
nonstructural functions compared to other techniques. However, there may still be some 
architectural and M/E/P elements, such as partitions and pipes adjacent to the columns, that 
require temporary removal. See Section 8.4.1 for discussions of fireproofing, asbestos, and 
concrete encasement. 

Removal of existing structural elements: To access the base plate, an existing slab and slab 
reinforcement will probably have to be removed and replaced. Care should be taken to not 
damage any of the existing structural elements. If the existing base plate is being replaced 
entirely, a column shoring scheme has to be devised. The existing anchor rods should be cut at 
the top of footing and the new rods have to be relocated. 

Construction loads: See general discussion in Section 8.4.1. Construction loads at the basement 
level are not typically a concern if a slab-on-grade is present. 

Proprietary Concerns 

Many grout and adhesive products are available. 

8.4.6 Enhance Beam-Column Moment Connection 
Deficiency Addressed by Rehabilitation Technique 

Riveted, bolted, and WSMF connections are upgraded to improve their ability to withstand 
inelastic rotational demands and develop the plastic moment capacity of the beams. 

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Description of the Rehabilitation Technique 

The techniques discussed in this section were developed specifically to address pre-Northridge 
WSMF connections. These techniques can be adapted with prudence to existing riveted and 
bolted connections that are found to be inadequate or to upgrade partially restrained connections. 
These techniques are covered thoroughly in AISC Design Guide 12 and FEMA 351 and only 
briefly presented here. The reduced beam section (RBS) is the only technique that weakens the 
beam in flexure, which in turn, moves the plastic hinge away from the column and reduces the 
demand on the complete joint penetration (CJP) welds. Two other methods - welded haunch and 
bolted bracket  also move the hinge away from the column but strengthen the existing 
connection and seek to maintain the original flexural capacity of the beam. A similar method 
employs cover plates over the beam flanges, requiring little additional space. This method is not 
discussed in detail here but more information can be found in FEMA 351. Additional 
modifications that should be performed for each technique include adding or verifying the 
capacity of beam flange continuity plates across the column web and strengthening the panel 
zone; see the references mentioned above and AISC Design Guide 13 (Carter, 1999) for 
reference. 

The selection of a particular connection modification depends on specific project factors. There 
are advantages and disadvantages to each of the methods. While each of the three connections 
discussed below consistently developed a minimum plastic rotation of 0.02 radian in cyclic 
loading experiments, the welded haunch and bolted bracket demonstrated higher levels of 
performance and reliability (Gross et al., 1999). On the other hand, the RBS modification 
requires no additional space and reduces the beam capacity to enforce a strong-column weak-
beam condition. The welded haunch is the only modification that exhibited desirable behavior in 
tests where the beam top flanges were left in a pre-Northridge condition (Gross et al., 1999). 
The bolted bracket scheme requires top flange reinforcement but eliminates field welding. 

Design Considerations 

Connection modifications merely ensure that the connections behave as originally intended in a 
moment frame. These modifications by themselves do not reduce the frame forces in an 
earthquake. Other structural upgrades are necessary if the frame does not have sufficient global 
strength or stiffness to resist the demands. 

Research basis: Numerous experimental programs have been performed under the auspices of 
the SAC Joint Venture, the National Science Foundation, the National Institutes of Standards and 
Technology, and AISC. Most of the results of these tests have been summarized in AISC Design 
Guide 12 and the FEMA documents listed in the references. The user is cautioned to extrapolate 
beyond the conditions that have been tested only with proper considerations of the differences 
between the tested connections and the building conditions. 

Design forces: AISC enforces the strong-column weak-beam concept by specifying a minimum 
column-beam moment ratio in the AISC Seismic Provisions. The general intent is to prevent 
plastic hinges from developing in the columns, which could lead to a soft story. Some yielding 
may still occur in the columns without causing loss in frame strengths. The required flexural 
strength of a column is determined from several variables associated with the beam. The beam 
plastic moment at the critical plastic section  centerline of RBS or tip of haunch and bracket  

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includes a strain hardening factor and the expected yield stress of the flanges. An additional 
moment is a function of the shear in the beam and the distance from the critical plastic section to 
the column centerline. This shear is a combination of the shear associated with the plastic 
moment and the gravity loads. 

RBS: The design of moment frames is often governed by global and interstory drift limits instead 
of strength requirements. Thus, the reduction in beam strength and stiffness from an RBS could 
be acceptable within practical limits. If the presence of an existing slab poses construction and 
design issues, then the top flange may remain intact and only the bottom flange has to be cut; 
though minimal removal of the existing slab is still required to replace the existing top flange 
weld. Note that a bottom flange modification only achieves a minimal stress reduction and that 
both the top and bottom flange CJP welds still require replacement with high toughness weld 
metal. Various RBS cut shapes have been tested successfully, though the radius cut RBS 
minimizes stress concentrations. The critical dimension is the depth of the cut, which can also be 
expressed in terms of flange width reduction as a percentage. AISC recommends a maximum 
flange reduction of 50% as a practical limit for field modifications. To determine if a flange 
reduction is adequate, the maximum moment at the face of the column can be computed for the 
modified beam. If the ratio of this moment to the plastic moment of the beam exceeds 1.05, a 
bottom flange RBS modification is not recommended. Either a top flange RBS should also be 
provided or another method should be used. 

Welded haunch: A tapered haunch welded to a beam bottom flange changes the force transfer 
mechanism at the connection. Analytical and experimental studies have found that the beam 
shear transfers through the haunch flange (FEMA, 2000b). This, in turn, reduces the shear 
stresses in the flange welds and increases the depth of the column panel zone. The welded 
haunch is the only connection out of the three presented for this technique to reach plastic 
rotations on the order of 0.03 radian while allowing the top flange weld to remain in a pre-
Northridge condition in tests. Additional rotational capacity can be obtained by upgrading the 
top flange groove weld to a higher toughness, adding a cover plate, or adding another haunch, in 
order of increasing performance. The two latter modifications can be performed without 
replacing the top flange weld. 

Bolted bracket: Considered an alternative to the welded haunch, this option does not require any 
field welding. A portion of the flange force is transferred to the bracket and reduces the weld 
stresses. Several bolted bracket configurations are possible and can be applied to either beam 
flange or both flanges for large beams. Common types of brackets include the haunch bracket, 
pipe bracket, angle bracket, and double angle bracket. Pipe and angle brackets are relatively 
compact and can be covered up by the slab for a top flange application. For bottom flange 
haunch only schemes, a stiff angle is still recommended at the top flange (Gross et al., 1999). 
The pre-Northridge weld in the top flange can remain in this case even if it is found to have 
defects. 

Material strength: Yield strengths of existing materials are best determined from tests of samples 
taken from the actual members. Samples taken from flanges are preferred over the web. When 
samples are not available, AISC Design Guide 12 prescribes some overstrength values for 
different grades of steel. 

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Minor axis connections: Moment connections to the minor axis of a column are not as common, 
but sometimes present in older moment frame construction. Their performance in the Northridge 
earthquake did not result in significant damage. Thus, little work has been done to study the 
rehabilitation of these connections. Similar concepts presented in this section may be used to 
rehabilitate these connections with the recognition that the columns may be weaker than the 
beams, which would also limit the inelastic demands on the connections. Note the governing 
jurisdiction may require qualification through the testing program described in FEMA 351. 

Nonstructural issues: Welded haunches and bolted brackets are considerable in size. The 
recommended welded haunch and bolted bracket depths are approximately one third and one half 
times the beam depths, respectively. For a W36 beam used in a moment frame, this adds up to 
18 inches to the connection depth. This makes the option of adding components to only the 
bottom flange more attractive, though it could still interfere with the ceiling. At the top flange, 
the modification would certainly have to be hidden or incorporated into an architectural feature. 

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 
available, and conducting comprehensive field surveys, the following issues should be noted: 

Weld filler metal matching and overmatching: Weld filler metals with greater tensile strength 
than the connected steel should be used. Flux cored arc welding and shielded metal arc welding 
electrodes that conform to E70 specifications exhibit overmatching properties compared to 
common steel specifications, including ASTM A36, A572 (Grade 42 and 50), A913 (Grade 50), 
and A992 (FEMA, 2000b). 

Weld metal toughness: Tests on existing connections modified with RBS but no modifications to 
the existing beam flange welds showed poor performance (Gross et al., 1999). Improved 
connection performance was achieved for connections where the top and bottom flange welds 
were replaced with higher toughness weld metal. AISC Seismic Provisions now require weld 
metals with minimum Charpy V-Notch (CVN) toughness of 20 ft-lbs at -20F and 40 ft-lbs at 
70F for demand critical welds. 

Weld backing and access holes: Backing plates provided for flange welds create a notch effect 
and also hinder detection of weld flaws at the weld root. Backing should be removed at flanges 
receiving new CJP groove welds and reinforced with fillet welds. The removal of existing 
backing without weld replacement is an ineffective upgrade since the existing weld is still likely 
to have low toughness. The size and shape of the weld access hole should be configured to 
provide welder access and minimize stress concentrations. FEMA 351 provides a specific weld 
access hole detail that meets these criteria. 

RBS (Figure 8.4.6-1): The distance from the face of column to the start of the RBS cut and the 
length of the cut, more concisely known as the offset and the chord length, respectively, should 
be kept small as to not allow the moment to increase significantly from the plastic hinge to the 
column. Yet, the distance to the start of the cut should also be large enough to allow for the 
flange force from the RBS to be uniformly distributed across the flange at the CJP weld. The 

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length of the cut should also be long enough to control the inelastic strains along the RBS. The 
radius of the cut is a function of its depth and the chord length. See AISC Design Guide 12 for 
specific recommendations and guidelines for selecting these dimensions. 

Figure 8.4.6-1: Reduced Beam Section at Bottom Flange of Existing Beam 

(adapted from AISC Design Guide 12) 

Welded haunch (Figure 8.4.6-2): Haunches can be cut from WT- or W-shapes. Typically, the 
haunch flange is groove welded to the beam and column flanges while the haunch web is fillet 
welded to these elements. Experimental programs have consistently used the same haunch 
geometrylength and depth approximately one half and one third that of the beam depth, 

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Techniques for the Seismic Rehabilitation of 
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Figure 8.4.6-2: Welded Haunch at Bottom Flange of Existing Beam 

(adapted from AISC Design Guide 12) 

respectively. The haunch flange is sized to resist most of the shear force from the beam. The 
haunch web size can be established to achieve equilibrium. At the intersection of the beam and 
haunch flanges, web stiffeners should be provided for the beam. Similarly at the column, add 
continuity plates to align with the haunch flange. 

Bolted bracket: The preferred configuration for this modification is a haunch bracket at the 
bottom flange and an angle bracket at the top flange that can be hidden within the slab. If 
additional reinforcement is necessary at the top flange, angles can be added below the top flange 

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on each side of the web, as shown in Figure 8.4.6-3. Close-fit holes minimize slip in this 
connection. Tests have demonstrated that these connections are essentially fully rigid and can 
reach large plastic rotations while allowing the pre-Northridge welds at the flanges to remain 
(Gross et al., 1999). AISC recommends neglecting the existing welds and designing each 
bracket for the entire flange tension force. The bottom flange haunch bracket will be slightly 
larger than the welded haunch for the same beam size. The top flange angle bracket may require 
several rows of bolts along the beam flange and has to be designed for the prying force on the 
column flange. Due to limitations in the available L-shapes, the angle bracket may have to be 
cut from a W-shape. 

Cost/Disruption 

Connection modifications are locally very disruptive. Modification of a connection typically 
requires access from two floors to perform the work on each flange. Noise associated with this 
type of work will spread and disrupt tenants on other floors unless the work is done during off-
hours. The RBS is probably cheaper than the other two modifications since it requires the least 
amount of material and labor. With older buildings, there may be asbestos present in the 
fireproofing around the steel members, which could require costly abatements to expose the 
connections. 

Construction Considerations 

The engineers involvement during the construction phase is critical during a seismic 
rehabilitation. The design of the retrofit scheme must not neglect the construction phase and 
should consider these issues at a minimum: 

Welding/bolting issues: See general discussion in Section 8.4.1. The bolted bracket modification 
can be performed without any field welding. Primary issues associated with the bolted bracket 
consist of typical field bolting issues such as set up, fit-up, and alignment. 

Removal of existing nonstructural elements: The connection work will affect ceilings, lights, and 
other mechanical/electrical/pluming components. Connection modifications at the roof level 
may warrant removal of the roofing and waterproofing. See Section 8.4.1 for discussions of 
fireproofing, asbestos, and concrete encasement. 

Removal of existing structural elements: See general discussion in Section 8.4.1. 

Construction loads: See general discussion in Section 8.4.1. 

Proprietary Concerns 

There are several proprietary connections that have been developed to upgrade pre-Northridge 
connections, generally sharing similar design intents. Some of these connections briefly 
presented in FEMA 351 include the Side Plate connection system, the Slotted Web connection, 
and one particular type of the bolted bracket connection. The reader should contact the licensors 
of these technologies for more information. 

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8.4.7 Enhance Column Splice 
Deficiency Addressed by Rehabilitation Technique 

Strengthen welded or bolted column splices that do not meet the detailing and minimum design 
strength requirements in AISC Seismic Provisions. 

Description of the Rehabilitation Technique 

Numerous options exist for upgrading a column splice. The approach and the level of 
strengthening depend on the type of lateral system since special moment frames have different 
requirements from other systems in the AISC Seismic Provisions. It also depends on if the 
existing splice is welded or bolted and if field welding is permitted. Field welding may be a 
necessity in some cases since the AISC Seismic Provisions do not allow bolts and welds to share 
loads on the same faying surface. 

Most existing welded splices are likely to be complete joint penetration (CJP) welded or partial 
joint penetration (PP) welded. CJP welds require beveled transitions to avoid stress 
concentrations. A beveled transition can be constructed in the field by building up the weld over 
the thicker column flange or grinding away a portion of the flange. PP welds inherently possess 
stress concentrations at the unwelded portion of the joint and thus, have higher strength 
requirements, but do not require beveled transitions. PP welds can be strengthened by welding 
plates or stiffeners across the splice. When field welding is impractical or undesirable, it may be 
possible to use bolted plates at an existing welded splice if the bolts are designed to resist the 
entire load. 

The rehabilitation method for a bolted splice depends on the controlling design mode for the 
splice  whether the splice is governed by net section fracture of the column, yielding of the 
bolts, or gross yield or net section fracture of the splice plate. Bolts that govern the existing 
splice capacity could be replaced with stronger bolts, such as replacing A325 with A490. The 
upgrade may also be as simple as replacing bolts that have threads included in the shear plane 
with longer bolts. Alternatively, more bolts or larger bolts could be provided, but both would 
require more extensive field work. Splice plates with insufficient strength can be replaced or 
new plates could be provided on the opposite side of the existing plate, which would require 
longer bolts. The existing plates may also be extended by welding new plates to the ends. 

Design Considerations 

Considerations for a column splice include the justification of a load path, preservation of 
symmetry to avoid eccentric loads, and deformation compatibility if welds and bolts are both 
used. In addition to the requirements in the AISC Seismic Provisions, the AISC Steel 
Construction Manual (AISC, 2005c) has a general discussion of column splices and contains 
typical details of W-shape splices. Information can be found regarding filler plate sizes for 
different column depths and erection tolerances. Important design issues include the following: 

Research basis: No references directly addressing the upgrade of existing column splices have 
been identified. 

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Design strength: The AISC Seismic Provisions specify a minimum strength requirement for all 
splices as a function of the expected yield strength and the flange area of the column. In 
addition, PP welded splices have higher strength requirements due to the stress concentration 
manifested in the crack-like notch at the unwelded side of a flange. Lastly, column splices in 
special moment frames have to be designed for the full flexural strength of the smaller column 
and their web splices require a shear strength of twice the plastic flexural strength of the column 
divided by the story height. 

PP welded splices: For flange thickness differences between the columns up to 1/8, steel shims 
can be fillet welded to the inside of the flanges. Greater than 1/8, it may be more practical to 
attach a plate to the flanges using CJP welds. 

Plate locations: Plates can be extremely versatile whether they are welded or bolted to the 
columns. Bolted plates are commonly placed at the outside of flanges, where extra filler plates 
can be used to make up differences in flange thicknesses. Filler plates are not necessary at the 
inside of flanges, where a flush surface is already provided if the same nominal depth columns 
are present. Plates can also be welded from one flange tip to another, which provides increased 
shear strength and stability across the joint. See Figure 8.4.7-1. 

Nonstructural issues: The column splices hardly take up any additional space but minor 
modifications and temporary relocations may be required for some architectural and M/E/P 
elements. 

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 
available, and conducting comprehensive field surveys, the following issues should be noted: 

Welded splices: Large shrinkage strains could develop when welding heavy sections. Bolted or 
fillet welded plate slices should be considered as an alternative if there is a possibility of a brittle 
weld fracture. 

Bolted splices: If new holes are drilled in the existing column or splice plate, tolerances of the 
existing holes should be verified to ensure that bolts will be loaded evenly. The capacity of the 
net section should also be checked. See Figure 8.4.7-2 for a sample detail. 

Web splices: Column webs in moment frames that use bolted web splices require plates or 
channels on both sides to reduce eccentricities. This should be considered for column webs in 
other lateral force-resisting systems though it is not a specific requirement in the AISC Seismic 
Provisions. 

Cost/Disruption 

The cost of implementing this technique is highly dependent on the level of modification 
performed to the existing splice. The level of disruption is typical for that of steel frame 
upgrades, involving vacancy of space, removal of all nonstructural elements around the column, 
and significant noise. 

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Techniques for the Seismic Rehabilitation of 
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Figure 8.4.7-1: Welded Splice Upgrade at Existing Column 

Construction Considerations 

The engineers involvement during the construction phase is critical during a seismic 
rehabilitation. The design of the retrofit scheme must not neglect the construction phase and 
should consider these issues at a minimum: 

Welding issues: See general discussion in Section 8.4.1. Though weld fractures were not found 
at column splices after the Northridge earthquake, an environment in which the welder can 
perform quality welds is critical. 

Removal of existing nonstructural elements: See general discussion in Section 8.4.1. Depending 
on the location of the splice, the upgrade could affect ceilings, lights, and other 
mechanical/electrical/pluming components. 

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Techniques for the Seismic Rehabilitation of 
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Figure 8.4.7-2: Bolted Splice Upgrade at Existing Column 

Removal of existing structural elements: Due to the critical nature of columns, the removal of 
existing welds or bolts at a column should be minimized. Column alignment and stability should 
be maintained at all times. 

Construction loads: See general discussion in Section 8.4.1. Typically, welding on a loaded 
column should not create a safety issue, although stability during construction should always be 
considered. At a minimum, see section in AISC Steel Construction Manual (2005c) on column 
splices and the AISC Code of Standard Practice for Steel Buildings and Bridges (AISC, 2005a) 
for other construction considerations. 

Proprietary Concerns 

There are no known proprietary concerns with this technique. 

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8.4.8 Add Steel Plate Shear Wall (Connected to An Existing Steel Frame) 
Deficiency Addressed by Rehabilitation Technique 

Moment frames buildings that are insufficient to resist lateral forces or too flexible to control 
building drifts can be strengthened and stiffened by adding steel plate shear walls (SPSW), as 
shown in Figure 8.4.8-1A. The shear walls may be used alone as the new lateral force-resisting 
system or in conjunction with the moment frames. 

Description of the Rehabilitation Technique 

Shear walls can add considerable strength and stiffness to a structure. It should be considered as 
an alternative to adding braced frames if additional stiffness is required. Compared to concrete 
shear walls, steel plate shear walls are lighter and add less seismic mass to a structure. They also 
take up less space and may be more economical to construct, particularly in taller buildings 
where the costs of delivering formwork and pumping concrete are significant. The behavior of 
this system is analogous to braced frames that rely on tension-only braces, as well as plate 
girders whereby tension fields develop along the diagonals. However, neither of these examples 
completely characterizes the behavior of an SPSW. The beams and columns in an SPSW behave 
as boundary elements that are subject to a complex array of forces and require a considerable 
amount of stiffness to develop the capacity of the steel panels. Large strut forces are imposed on 
the beams while columns are subject to axial, flexural, and shear forces. Thus, preference of 
shear wall locations should be given to existing bays of moment frames, in which the members 
and connections are less likely to require modifications for use in the SPSW system. The energy 
dissipating mechanisms in an earthquake include tension yielding and eventual tearing of the 
diagonal tension fields, shear yielding of the plates, compressive buckling of the plates along the 
diagonal compression fields, and slipping of the bolted connections to the fin plates when used. 

In an existing building, fin plates at the wall boundaries would be field welded to the beams and 
columns first. Large steel panels would then be welded or bolted to these fin plates, as illustrated 
in Figure 8.4.8-1B. Panel splices could also be welded or bolted in the field. If welded splices 
are used, it is recommended to provide full penetration welds with the backing bars removed 
after welding. Openings are acceptable if stiffeners are provided around the edges. The panels 
themselves may be unstiffened or stiffened. Provisions for the design of this system have been 
included in the 2005 AISC Seismic Provisions. AISC is also in the process of publishing a 
design guide for unstiffened SPSW. 

Design Considerations 

Adding shear walls to a moment frame building increases its stiffness considerably. The 
upgraded structure should be evaluated for higher lateral and overturning forces accordingly. 
This system is almost always controlled by shear due to the substantial flexural strength provided 
by the steel column boundary elements. 

Research basis: No references directly addressing the addition of SPSW to moment frame 
buildings have been identified. However, if the inelastic deformations can be limited to the steel 
plates, the seismic performance of the strengthened structure is not expected to differ from new 

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Techniques for the Seismic Rehabilitation of 
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Figure 8.4.8-1A: Unstiffened Steel Plate Shear Wall 

buildings utilizing SPSW. A summary of the major research on this system as well as its seismic 
performance in past earthquakes can be found in Astaneh-Asl (2001). 

Unstiffened vs. stiffened walls: Stiffened walls tend to exhibit higher shear strength though both 
types of walls can be expected to exhibit ductile behavior. Buckling of the steel plates in 
unstiffened walls allow tension fields to develop and resist the lateral forces. Stiffeners, such as 
plates or channels, can be welded to the steel panels to prevent buckling of steel plates. These 
walls are more likely to yield in shear instead of developing tension fields. The use of stiffeners 
also permits the panels themselves to be thinner than panels in unstiffened walls. The panels in 
stiffened wall participate in resisting the overturning forces because buckling does not occur in 
the panels. Consequently, overturning forces in unstiffened walls are primarily resisted by the 
columns. 

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Figure 8.4.8-1B: Fin Plate Connection Options 

Nonstructural issues: The addition of walls to an existing structure changes the architectural 
character of the building. Walls at the exterior will be visible in buildings with clear glazing. At 
interior bays, walls have to be configured to avoid obstruction of existing corridors, doorways, 
and other building systems. Walls can be exposed and incorporated into the interior architecture 
or hidden in partition walls. 

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Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 
available, and conducting comprehensive field surveys, the following issues should be noted: 

Steel connections: Welds from the fin plates to beams and columns should develop the capacity 
of the steel panels through full penetration welds or fillet welds on both sides of the plates. To 
allow for construction and field tolerances, the steel panels can be lapped with the fin plates and 
connected using fillet welds along both edges of the lap. 

Connections at slabs: To avoid damaging the slab reinforcing steel critical for transferring 
diaphragm forces to the shear walls, the concrete slab could be chipped away without damaging 
the slab steel and individual fin plates could be placed between the reinforcing steel. The 
diaphragm forces are transferred to the wall through shear studs on the beams. Alternatively, if 
continuous fin plates are required and/or the shear studs are inadequate, the reinforcing steel in 
could be cut and welded directly to the fin plates. In this case, forces transfer from the slab into 
the wall through shear friction. 

Cost/Disruption 

As always, the cost of a strengthening scheme depends on the project and its unique 
requirements. There are no issues with the SPSW system that is known to cost significantly 
more than adding braced frames or concrete shear walls to a moment frame building. 
Unstiffened walls are cheaper and less labor intensive than stiffened walls. New foundations are 
almost always required for new walls and could be extremely costly if deep foundations, such as 
drilled piers, are added. 

Installing new walls is disruptive to the occupants because of the noise and vibrations associated 
with construction. Even if tenants are relocated to parts of the building where the work is not 
being performed, vibrations associated with cutting, chipping, and drilling of concrete as well as 
the installation of steel panels can transmit through the structure. The disruption can be reduced 
somewhat if the walls are installed at the perimeter. 

Construction Considerations 

See Section 8.4.1 for general discussions of welding issues, removal of existing nonstructural 
and structural elements, and construction loads. 

Proprietary Concerns 

There are no known proprietary concerns with this technique. 

8.4.9 Convert an Existing Steel Gravity Frame to a Moment Frame 
Deficiency Addressed by Rehabilitation Technique 

This technique addresses moment frames buildings that only require slight gains in strength 
and/or stiffness. The extent of connection modification depends on the seismic hazard and the 
type of moment frame. 

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Description of the Rehabilitation Technique 

Converting existing gravity frame connections to moment frame connections does not increase 
the strength or stiffness of a structure significantly unless a large majority of the gravity frame 
column connections are made moment resisting. The strength and stiffness gain are also limited 
because the existing beams and columns used in the gravity frame are typically much lighter than 
the moment frame members. However, these members could also be strengthened as part of the 
rehabilitation scheme. This technique has less impact on the architectural character of the 
building than adding braced frames or shear walls. 

The simplest method for implementing this technique is through the addition of welded flange 
cover plates from the beam to the column without any modification of the bolted shear tab. The 
beam flanges remain unattached from the column since the gap typically exceeds the maximum 
permitted root opening size for full penetration welds. This method should only be used with 
Ordinary Moment Frame applications. 

In high seismic regions where moment frames have to meet Special Moment Frame 
requirements, a more sophisticated connection upgrade that forces the beam yielding to occur 
away from the connection should be provided. This could range from adding welds to the bolted 
shear tab for the method described above to welding top and bottom haunches from the beam to 
the column. Several methods are presented in Section 8.4.6 as well as FEMA 350 (FEMA, 
2000a) and FEMA 351 (FEMA, 2000b). The level of upgrade ultimately depends on the 
expected ductility demand and the performance objective. 

Design and Detailing Considerations 

The strong-column weak-beam concept should still be a primary design consideration. All 
detailing issues related to the design of moment frame connections need to be considered, 
including weld filler metal matching, weld metal toughness, removal of weld backing, and 
column flange and web reinforcing. For Ordinary Moment Frames, joint reinforcing should be 
limited in order to permit some yielding, as to not place excessive demands on the flange to 
column connections. For Special Moment Resisting Frames, other requirements such as width-
thickness limitations, lateral bracing requirements, etc., should be checked to be in accordance 
with the AISC Seismic Provisions. In most cases, it is expected that the existing gravity beam 
and column configurations will be such that it will be difficult to meet the requirements for 
Special Moment Frames without other major modifications. As a result, it is expected that in 
most cases, the Ordinary Moment Frame requirements would apply. 

Cost/Disruption 

These issues are discussed in Section 8.4.6 for moment connections. Connection upgrades are 
typically less disruptive than adding braced frames or shear walls. The costs could vary 
depending on if existing moment frame connections are also being upgraded and the total 
number of connections being modified. 

Construction Considerations 

See Section 8.4.6 for general discussions of issues related to moment connections. 

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8.5 References 
ACI, 1991, Guide to Certification of Shotcrete Nozzlemen, ACI-506.3R, American Concrete 
Institute, Farmington Hills, MI. 

ACI, 1994, Guide for the Evaluation of Shotcrete, ACI-506.4R, American Concrete Institute, 
Farmington Hills, MI. 

ACI, 1995a, Guide to Shotcrete, ACI-506R, American Concrete Institute, Farmington Hills, 
Michigan. 

ACI, 1995b, Specification for Shotcrete, ACI-506.2, American Concrete Institute, Farmington 
Hills, MI. 

ACI, 1998, State of the Art Report on Fiber Reinforced Shotcrete, ACI-506.1R, American 
Concrete Institute, Farmington Hills, MI. 

ACI, 2005, Building Code Requirements for Reinforced Concrete and Commentary, ACI 318, 
American Concrete Institute, Farmington Hills, MI. 

AISC, 2005a, Code of Standard Practice for Steel Buildings and Bridges, American Institute of 
Steel Construction, Chicago, IL. 

AISC, 2005b, Seismic Provisions for Structural Steel Buildings, American Institute of Steel 
Construction, Chicago, IL. 

AISC, 2005c, Steel Construction Manual, 13th Edition, American Institute of Steel Construction, 
Chicago, IL. 

Astaneh-Asl, A., 2001, Seismic Behavior and Design of Steel Shear Walls, Structural Steel 
Educational Council, Moraga, CA. 

Carter, C.J., 1999, AISC Design Guide 13, Stiffening of Wide-Flange Columns at Moment 
Connections: Wind and Seismic Applications, American Institute of Steel Construction, Chicago, 
IL. 

FEMA, 2000a, Recommended Seismic Design Criteria for New Steel Moment-Frame Buildings, 
FEMA 350, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 2000b, Recommended Seismic Evaluation and Upgrade Criteria for Existing Steel 
Moment-Frame Buildings, FEMA 351, Federal Emergency Management Agency, Washington, 

D.C. 
FEMA, 2000c, State of the Art Report on Connection Performance, FEMA 355D, Federal 
Emergency Management Agency, Washington, D.C. 

FEMA, 2000d, State of the Art Report on Past Performance of Steel Moment-Frame 

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Techniques for the Seismic Rehabilitation of 
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Buildings in Earthquakes, FEMA 355E, Federal Emergency Management Agency, Washington, 

D.C. 
FEMA, 2000e, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 
356, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 2003, NEHRP Recommended Provisions for Seismic Regulations for New Buildings and 
Other Structures, FEMA 450, Federal Emergency Management Agency, Washington, D.C. 

Fisher, J.M. and L.A. Kloiber, 2006, AISC Design Guide 1, Base Plate and Anchor Rod Design, 
American Institute of Steel Construction, Chicago, IL. 

Griffis, L.G., 1992, AISC Design Guide 6, Load and Resistance Factor Design of W-Shapes 
Encased in Concrete, American Institute of Steel Construction, Chicago, IL. 

Gross, J.L., Engelhardt, M.D., Uang, C.M., Kasai, K., and N.R. Iwankiw, 1999, AISC Design 
Guide 12, Modification of Existing Welded Steel Moment Frame Connections for Seismic 
Resistance, American Institute of Steel Construction, Chicago, IL. 

Lee, H. and S.C. Goel, 1990, Seismic Behavior of Steel Built-up Box-Shaped Bracing 
Members, and Their Use in Strengthening Reinforced Concrete Frames, Report No. UMCE 907, 
University of Michigan, Ann Arbor, MI. 

Uriz, P. and S. Mahin, 2004, Summary of Test Results for UC Berkeley Special Concentric 
Braced Frame No. 1 (SCBF-1) Draft, Version 1.1, Department of Civil and Environmental 
Engineering, University of California, Berkeley, CA. 

Yang, F. and S. Mahin, 2005, Limiting Net Section Fractures in Slotted Tube Braces, Structural 
Steel Educational Council, Moraga, CA. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames 

Chapter 9 - Building Types S2/S2A: Steel Braced Frames 

9.1 Description of the Model Building Type 
Building Type S2 consists of a frame assembly of steel beams and columns. Lateral forces are 
resisted by diagonal steel members placed in selected bays. Floors are cast-in-place concrete 
slabs or concrete fill over metal deck. These buildings are typically used for buildings similar to 
steel moment frames, although more often for low-rise applications. Figure 9.1-1 shows an 
example of this building type. 

Building Type S2A is similar but has floors and roof that act as flexible diaphragms such as 
wood or untopped metal deck. This is a relatively uncommon building type and is used 
primarily for small office or commercial buildings in which the fire rating of concrete floor is not 
needed. 

Figure 9.1-1: Building Type S2: Steel Braced Frames 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames 

Variations Within the Building Type 

The two principal types of braced frame configurations are concentrically braced frames (CBFs) 
and eccentrically braced frames (EBFs). In a CBF, the centerlines of members that meet at a 
joint all intersect at a single point. These frames behave as vertical truss systems by transferring 
lateral loads primarily through axial loading of beams, columns, and braces. During 
earthquakes, inelastic behavior is typically limited to the braces and connections. Common CBF 
configurations include diagonal bracing, X-bracing (or 2-story X-bracing), V-bracing (or 
inverted-V-bracing), and K-bracing. The diagonal braces in an EBF are offset at joints such that 
link beams separate the ends of braces from columns or other braces. Inelastic behavior is 
concentrated in the links while all members outside of the links remain elastic or near elastic. 
Link beams can be located adjacent to a column or at the center of a beam. Common types of 
braces include W-shapes, hollow structural sections (HSS), steel pipes, double angles, and 
double channels. 

Braces are either welded directly to the beams and columns or welded to gusset plates. It is 
standard practice on the West Coast to weld gusset plates to the beams and columns. Away from 
the West Coast, the plate is more commonly welded to the beam and bolted to the column with a 
pair of angles or a WT-shape. Beam-column connections vary depending on whether there is a 
brace at the joint or not. Connections range from simple shear tabs to fully welded moment 
connections. 

Floor and Roof Diaphragms 

Diaphragms associated with this building type may be either rigid or flexible. The typical rigid 
diaphragm found in modern buildings consists of structural concrete on metal deck. Diaphragm 
forces transfer to the frames through shear studs welded to the beams. Older steel buildings that 
were constructed before metal decks were commonly used may have concrete slabs or masonry 
arches that span between the beams. Flexible diaphragms include bare metal deck or metal deck 
with nonstructural fill. These are frequently used on roofs that support light gravity loads. 
Decks could be connected to the steel members with shear studs, puddle welds, screws, or shot 
pins. The steel members also act as chords and collectors for the diaphragm. 

Foundations 

There is no typical foundation for this building type. Foundations can be of any type, including 
spread footings, mat footings, and piles, depending on the characteristics of the building, the 
lateral forces, and the site soil. Spread footings are used when lateral forces are not very high 
and a firm soil exists. For larger forces and/or poor soil conditions, a mat footing below the 
entire structure is commonly used. Pile foundations are used when lateral forces are extremely 
large or poor soil is encountered. The piles can be either driven or cast-in-place. Vertical forces 
are distributed to the underlying soil through a combination of skin friction between the pile and 
soil and/or direct bearing at the end of the pile; lateral forces are resisted primarily through 
passive pressure on the vertical surfaces of the pile cap and piles. 

9.2 Seismic Performance Characteristics 
Braced frames are generally considered to be stiff systems in the elastic range. Their nonlinear 
response depends on their ability to redistribute forces between bays and drifts between stories. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames 

Braces and connections in CBF undergo large inelastic deformations in tension and compression 
into the post-buckling range. Ductility of CBF systems in past earthquakes have been limited by 
local failures of braces and connections. It is thought that CBF systems that are properly 
designed and detailed can possess ductility in excess of that previously assigned to these systems 
(AISC, 2005). Yet, recent experimental testing at UC Berkeley found that special concentrically 
braced frames (SCBF), designed with an inverted-V configuration using the 1997 AISC Seismic 
Provisions, fractured the HSS braces after only a few cycles of loading in the inelastic range 
(Uriz and Mahin, 2004). Also, the damage concentrated at the level that first experienced brace 
buckling, resulting in weak story response. As of this writing, these results are still under 
investigation. 

EBF systems approach the higher performance levels of structural systems such as buckling-
restrained braced frames and fluid or friction damped frames (Horne et al., 2001). By limiting 
nonlinear action to the link beams, the post-yield behavior of the system and the maximum 
demand on the frame can be better predicted. The ductility of an EBF system is dependent on 
the length and detailing of the link beams and whether shear or flexural yielding governs their 
inelastic response. 

9.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
Undesirable behaviors of CBF systems that have been observed in past earthquakes include 
fracture of connection elements, fracture of braces, and local buckling of braces. These failures, 
in turn, cause excessive demands on other elements in the system and lead to overall frame 
failure. See Table 9.3-1 for deficiencies and potential rehabilitation techniques particular to this 
system. Selected deficiencies are further discussed below by category. 

Global Strength 

The lack of global strength to resist the seismic demands is a direct result of weak frames. In 
some cases, the existing beams and columns in a frame may possess enough capacity to 
accommodate upgrades to the braces and connections. If the beams and columns cannot make 
this accommodation, braces can be added to other bays to create new frames and thereby, reduce 
the demand on the existing frames. This would be more easily achieved for CBF systems than 
for EBF systems, which have special link beam requirements (AISC, 2005). 

Global Stiffness 

Braced frames are extremely stiff in the elastic range. Concerns with global stiffness occur in 
the inelastic range when braces are prone to buckling and cause a loss of stiffness. Limited post-
elastic stiffness is potentially provided by the frame and non-frame columns in the system (Tang 
and Goel, 1987; Hassan and Goel, 1991). 

Configuration 

Concentrically braced frames: K-bracing and inverted-V-bracing configurations exhibit 
undesirable behavior when a brace buckles. The remaining tension brace at a joint imparts an 
unbalanced force onto the column or beam. This is particularly hazardous for frames with K-
bracing since it may cause a column to fail entirely. In frames that use inverted-V-bracing, the 
unbalanced force is additive to gravity loads on beams. 

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Techniques for the Seismic Rehabilitation of 
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Table 9.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S2/S2A Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce Demand Remove Selected 
Components 
Global 
Strength 
Insufficient 
frame strength 
Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Steel plate shear wall 
[8.4.8] 
Strengthen braces 
[9.4.2], beams [8.4.3], 
columns [8.4.3], and/or 
connections [9.4.1] 
Seismic isolation 
[24.3] 
Supplemental 
damping [24.4] 
Global 
Stiffness 
Excessive drift Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Steel plate shear wall 
[8.4.8] 
Strengthen braces 
[9.4.2], beams [8.4.3], 
columns [8.4.3], and/or 
connections [9.4.1] 
Supplemental 
damping [24.4] 
Configuration Soft story Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Steel plate shear wall 
[8.4.8] 
Re-entrant 
corner 
Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Collector [8.4.4] 
Enhance detailing 
[8.4.3], [8.4.4] 
Load Path Missing 
collector 
Collector [8.4.4] 
Inadequate 
shear, flexural, 
and uplift 
anchorage to 
foundation 
Embed column into a 
pedestal bonded to other 
existing foundation 
elements [8.4.5] 
Provide steel shear 
lugs or anchor bolts 
from base plate to 
foundation [8.4.5] 
Inadequate out-
of-plane 
anchorage at 
walls connected 
to diaphragm 
Tension anchors 
[16.4.1] 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames 

Table 9.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S2/S2A Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce Demand Remove Selected 
Components 
Component 
Detailing 
Inadequate 
capacity of 
braces and/or 
connection 
Replace braces [9.4.1] Increase area of braces 
[9.4.2] 
Make braces composite 
elements [9.4.2] 
Improve b/t ratios 
[9.4.2] 
Add bolts and welds 
[9.4.1] 
Increase size of 
gusset plates [9.4.1] 
Inadequate 
capacity of 
beams, 
columns, and/or 
connections 
Add cover plates or box 
members [8.4.3] 
Provide gusset plates or 
knee braces [9.4.1] 
Provide gusset 
plates [9.4.1] 
EBFs not 
conforming to 
current 
standards 
Check current EBF 
design standards [9.4.2] 
Inadequate 
capacity of 
horizontal steel 
bracing 
Provide additional 
secondary bracing 
[9.4.2] 
Strengthen bracing 
elements [9.4.2] 
Reduce unbraced 
lengths [9.4.2] 
Strengthen 
connections [9.4.1] 
Diaphragms Inadequate 
in-plane 
strength and/or 
stiffness 
Collectors to distribute 
forces [8.4.4] 
Moment frame 
Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Concrete topping slab 
overlay 
Wood structural panel 
overlay at flexible 
diaphragms [22.2.1] 
Strengthen chords 
[8.4.3], [8.4.4], and 
[22.2.2] 
Add nails at flexible 
diaphragms [22.2.1] 
Inadequate 
shear transfer to 
frames 
Provide additional 
shear studs, 
anchors, or welds 
[22.2.7] 
Inadequate 
chord capacity 
Add steel members or 
reinforcement [8.4.3], 
[8.4.4] 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames 

Table 9.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S2/S2A Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce Demand Remove Selected 
Components 
Diaphragms 
(continued) 
Excessive 
stresses at 
openings and 
irregularities 
Add reinforcement 
[8.4.3] 
Provide drags into 
surrounding diaphragm 
[8.4.4] 
Infill opening 
[22.2.4], 
[22.2.6] 
Foundations See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames 

Eccentrically braced frames: The inelastic response of a link beam is influenced by its length 
relative to the ratio MP/Vp of the link section (AISC, 2005). Link beams that exhibit shear 
yielding have greater inelastic deformation capacity than ones that exhibit flexural yielding. 
AISC Seismic Provisions (AISC, 2005) permit shear yielding links to have four times the plastic 
rotation angles of flexural yielding links. Configurations with link beams adjacent to columns 
are susceptible to weld fractures found in pre-Northridge connections unless special detailing 
measures are taken to reduce the demands on these welds. 

Load Path 

Load path deficiencies in steel braced frame buildings include inadequate connections, 
collectors, and frame anchorage to foundations. Brace connections may have less strength than 
the braces. In Type S2 buildings, seismic forces transfer from the diaphragm to the frame 
through shear studs welded to collectors or directly to the frame beams. The collectors or the 
connections to the frame may be too weak to transfer these forces. Connections from columns to 
base plates or pile caps must resist shear, flexural, and potential uplift forces. Connections that 
cannot develop these frame forces prevent the frame from developing its full capacity. 

Component Detailing 

CBF: HSS and pipe braces with high b/t ratios and other steel shapes that lack compactness are 
subject to local buckling or fracture after a limited number of inelastic cycles. Ductility of these 
braces can be improved by infilling with concrete or adding longitudinal stiffeners. 
Alternatively, the braces can be replaced with double HSS sections, which can be used in 
configurations similar to double angles or channels (Lee and Goel, 1990). HSS and pipe braces 
are also subject to net section fracture at the gusset plate slots (Uriz and Mahin, 2004). This 
deficiency can be mitigated by adding reinforcing plates to the sides of the HSS without the 
slots, as shown in Figure 8.4.1-1 and Figure 8.4.1-2. For pipes, the reinforcing plates can be 
oriented at right angles to the pipe and appear like stiffeners. Brace connections that are only 
designed for the axial capacity of the braces may not be adequate to generate the full strength of 
the braces. To achieve good post-inelastic response, all eccentricities in the connection must be 
considered. A brace that buckles in the plane of the gusset plates should have its end 
connections designed for the full axial load and flexural strength of the brace (AISC, 2005). A 
brace that buckles out-of-plane should ensure that the gusset plates can develop restraint-free 
plastic rotations without buckling. 

EBF: Frames that rely on link-to-column connections have traditionally utilized similar detailing 
as pre-Northridge connections at beam-column joints. These connections should be reevaluated 
in the wake of the findings following the Northridge earthquake. Also, experimental research 
has found that link beams at the first floor undergo the largest inelastic deformation and have the 
potential to create a soft story (AISC, 2005). 

Diaphragm Deficiencies 

Common diaphragm deficiencies include insufficient in-plane shear strengths, inadequate 
chords, and excessive stresses at openings. Causes for these deficiencies could be due to lack of 
slab or fill thickness, lack of reinforcing steel in the slabs, insufficient connections to chord 
elements, and poor detailing at openings. See Chapter 22 for common rehabilitation techniques. 

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Techniques for the Seismic Rehabilitation of 
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Foundation Deficiencies 

Foundations that are inadequate do not develop the full capacity of the lateral force-resisting 
system. Their deficiencies result from insufficient strengths and sizes of footings, grade beams, 
pile caps, and piles. See Chapter 23 for common rehabilitation techniques. 

9.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
9.4.1 Enhance Braced Frame Connection 
Deficiency Addressed by Rehabilitation Technique 

Adequate capacities of connections are essential to the proper performance of a braced frame. 
Connections with insufficient strength and/or ductility to develop stable inelastic frame behavior 
are strengthened or replaced. 

Description of the Rehabilitation Technique 

Brace end connections commonly rely on additional connection elements (e.g., gusset plates), 
but the braces may also be directly welded to the beam and column through a moment 
connection. Moment connections are also used for link-to-column connections in an EBF and 
sometimes used for beam-to-column connections in both EBF and CBF systems. The mitigation 
approach is different depending on whether the existing connection relies on a connection 
member or a moment connection. 

If the existing connection members have sufficient capacity, the most economical alternative 
may be to increase the connection capacity by providing additional welds or bolts. This typically 
only allows for a limited increase in capacity since existing brace connection configurations can 
rarely accommodate significant modifications. If the existing connection members have 
inadequate capacity, the existing configuration and accessibility need to be assessed to determine 
whether adding supplemental connection members or replacing the existing connection members 
with members of greater capacity is more economical. Supplementing the existing connection 
eliminates the challenges associated with removal of existing connection welds and temporary 
support of the braces. 

The primary concern with moment connections used in braced frames is the use of low notch 
toughness weld metals. For braces that are expected to develop plastic hinges at their ends, 
consider replacing the existing welds with a high notch toughness weld metal. Beam-to-column 
connections in CBF do not typically experience large flexural forces and likely do not need to be 
upgraded. An exception occurs when the braces or their end connections fail and frame action 
becomes the primary mechanism for lateral force resistance; however, this is not a recommended 
design approach, as it does not ensure stable post-elastic behavior. In EBF configurations where 
the link beam is adjacent to the column, an upgrade should be considered given the large 
demands on the link beam and its critical nature. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames 

Design Considerations 

AISC Seismic Provisions does not permit the sharing of loads by both welds and bolts on the 
same faying surface. Thus, a bolted brace to gusset plate connection should only be enhanced 
with bolts or replaced entirely with welds. Note it is not uncommon in some regions for a gusset 
plate to be bolted to the column through a shear plate and welded to the beam since these are 
separate faying surfaces. In addition to having thorough knowledge of the existing material 
behaviors to ensure that the new and existing elements to interact in the desired manner, other 
design issues include the following: 

Research basis: No references directly addressing upgrades of braced frame connections have 
been identified. 

Design forces: Brace connections that are only designed for the axial capacity of the braces may 
not be adequate to generate the full strength of the braces. A brace that buckles in the plane of 
the gusset plates should have its end connections designed for the full axial load and flexural 
strength of the brace (Astaneh-Asl et al., 1986). This recommendation may be more appropriate 
for high seismic applications. 

Bolted connections: Bolts that govern the existing connection capacity could be replaced with 
stronger bolts, such as replacing A325 with A490. The upgrade may also be as simple as 
replacing bolts that have threads included in the shear plane with longer bolts. More bolts could 
be added if the existing configuration allows for it. Larger bolts could be provided but this 
would reduce the net section capacities due to enlargement of the holes. 

Welded connections: Existing fillet welds can be thickened provided the welds are not of low 
notch toughness weld metals or found to be inadequate through material testing. Otherwise, the 
existing welds should be removed and replaced with high notch toughness weld metals. A 
typical fully welded connection appropriate for use in SCBF is shown in Figure 8.4.1-1. For low 
to moderate seismic applications, Figure 8.4.1-2 shows a more compact connection. Welded 
brace connections to the weak axis of columns are complex and expensive; an example of this 
type of connection is shown in Figure 8.4.1-3. 

Moment connections: As mentioned above, moment connections that are subject to large flexural 
forces should be upgraded with high notch toughness weld metals. This would primarily apply 
to braces that directly connect to beams and columns in SCBF and some OCBF in high seismic 
applications. EBF link beams that are adjacent to columns also fall into this category. Link 
beams develop large flexural forces whether shear or flexural yielding governs. 

Nonstructural issues: Gusset plates designed in accordance with the AISC Seismic Provisions for 
SCBF can be fairly large and should be discussed with the architect and tenants. 

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 
available, and conducting comprehensive field surveys, the following issues should be noted: 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames 

Gusset plates: A brace in a SCBF that buckles out-of-plane could form plastic hinges at midspan 
and in the gusset plates at each end. The gusset plates should provide restraint-free plastic 
rotations without buckling. AISC Seismic Provisions suggests a minimum distance of two times 
the plate thickness between the end of the brace and the assumed line of restraint to achieve this. 
Consider increasing this distance to three times the plate thickness to accommodate over cutting 
of the slots in HSS and other erection tolerances. Connections that do not allow for restraint-free 
plastic rotations out-of-plane should be used primarily in situations where in-plane buckling of 
the braces govern or ductility demands are low. 

Bolted connections: If new holes are drilled in the existing brace and connection member, 
tolerances of the existing holes should be verified to ensure that bolts will be loaded evenly. 

Weld filler metal matching and overmatching: Weld filler metals with slightly greater tensile 
strength than the connected steel should be used. Flux cored arc welding and shielded metal arc 
welding electrodes that conform to E70 specifications exhibit overmatching properties compared 
to common steel specifications, including ASTM A36, A572 (Grade 42 and 50), A913 (Grade 
50), and A992 (FEMA, 2000b). 

Weld metal toughness: AISC Seismic Provisions now require weld metals with minimum Charpy 
V-Notch (CVN) toughness of 20 ft-lbf at -20F for all welds in the lateral force-resisting system. 
There is also an additional requirement of 40 ft-lbs at 70F for demand critical welds. 

Cost/Disruption 

Connection modifications are locally very disruptive. Noise associated with this type of work 
will spread and disrupt tenants on other floors unless the work is done during off-hours. These 
modifications could be particularly costly if existing gusset plates have to be replaced. 

Construction Considerations 

See Section 8.4.1 for general discussions of welding issues, removal of existing nonstructural 
and structural elements, and construction loads. 

Proprietary Concerns 

There are no known proprietary concerns with this technique. 

9.4.2 Enhance Strength and Ductility of Braced Frame Member 
Deficiency Addressed by Rehabilitation Technique 

Inadequate beams, columns, and braces are strengthened or replaced to achieve ductile frame 
behavior. 

Description of the Rehabilitation Technique 

The ductility of an existing brace can be enhanced by reducing its slenderness, which can be 
accomplished by decreasing its unbraced length, infilling hollow sections with concrete, or 
adding longitudinal stiffeners. The unbraced length of a brace can be reduced by adding 
secondary bracing members that are not part of the primary lateral force-resisting system. 
Infilling existing hollow sections with concrete can reduce the severity of local buckling (Liu and 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames 

Goel, 1988; Lee and Goel 1987). An effective width-thickness ratio for the infilled member is 
determined by multiplying the width-thickness ratio of the section by the factor (0.0082 x KL/r + 
0.264), applicable to braces with KL/r values between 35 and 90 (Goel and Lee, 1992). Adding 
longitudinal stiffeners presents the least field complications; the stiffeners could consist of plates 
or small angle sections. 

If both strength and ductility are required, new braces have to be added. Some configurations 
may lend themselves to schemes that allow the existing braces to remain. These include single 
angle, double angle, and channel braces that can be doubled; rolled sections can also be cover 
plated. In other cases, it is more practical to replace the existing brace with a new brace, of 
which numerous options exist. The increase in brace strength may require upgrades to other 
components of the braced frame, such as the brace connections, beams, and columns. 
Connection upgrades are discussed in the Section 9.4.1. In many cases, the most cost-effective 
alternative for increasing the capacity of the existing beams and columns in is to add cover plates 
or side plates to create box sections. This technique is discussed in Section 8.4.3. 

Design Considerations 

It would be preferable to limit the strengthening of the existing braces to the capacity of the other 
members of the lateral force-resisting system, including the foundations, to avoid triggering too 
many upgrades. Thorough knowledge of the existing material behaviors and strengths are 
necessary for the new and existing elements to interact in the desired manner. Other design 
issues include the following: 

Research basis: No references directly addressing upgrades of braced frame members have been 
identified. 

Existing brace strengthening: Significant modifications to an existing brace could trigger 
strengthening or redesign of its end connections. Strengthening of existing K- or inverted-Vbracing 
should be undertaken only after careful evaluation of the additional bending forces 
following the buckling of the compression bracing. Where the existing bracing in these systems 
is found to have inadequate capacity, the preferred solution is to replace it with a diagonal or X-
bracing configuration. 

Secondary bracing: A brace member is designed to resist both tension and compression forces, 
but its capacity for compression stresses is limited by potential buckling and is therefore less 
than the capacity for tensile stresses. Since the design of the system generally is based on the 
compression capacity of the brace, some additional capacity may be obtained by simply reducing 
the unsupported length of the brace by means of secondary bracing provided the connections 
have adequate reserve capacity or can be strengthened for the additional loads. 

New brace selection: If existing braces are replaced, use compact and non-slender sections 
whenever possible to avoid premature fracturing or buckling of the braces during post-yield 
behavior. Two particular brace types not common in older braced frame buildings are double 
HSS sections and buckling-restrained braces. Double HSS sections can be used in 
configurations similar to double angles or channels (Lee and Goel, 1990). They provide reduced 
fit-up issues and smaller width-to-thickness ratios compared to a single HSS, resulting in 

9-11 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames 

increased energy dissipation capacity. The other type of brace, used in a buckling-restrained 
braced frame (BRBF), is typically used in new buildings but has also been used successfully in 
new BRBF systems in existing buildings. One example of a brace used in a BRBF consists of a 
steel core inside a casing, which consists of a hollow structural section (HSS) infilled with 
concrete grout. Proprietary materials separate the steel core and concrete to prohibit bonding 
between the two materials. There are other buckling-restrained braces that do not use grout or 
additional separating agents between the steel and grout. The main advantage of these braces is 
the ability of the casing to restrain the buckling of the steel core without providing any additional 
axial force resistance beyond the capacity of the steel core. Provisions for new building BRBF 
design are included in the NEHRP Recommended Provisions (FEMA, 2003) and the AISC 
Seismic Provisions for Structural Steel Buildings. Note that significant connection modifications 
may be required when braces are replaced. 

Nonstructural issues: Brace modifications, when exposed, will affect the interior architecture or 
if hidden in partition walls, these walls may be thicker than typical walls. Beams that are 
increased in size affect nonstructural components by reducing clear floor heights. These 
components typically include suspended ceilings, pipes, conduits, and ducts. Coordination with 
the architect and other trades should not be overlooked or underestimated. 

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 
available, and conducting comprehensive field surveys, the following issues should be noted: 

Built-up brace members: While double angles, double channels, and double HSS offer 
advantages for installation, special criteria apply to these members when used in a SCBF. 
Buckling of these types of braces imposes large shear forces on the stitches. Therefore, closer 
stitch spacing and higher stitch strengths are required. More stringent member compactness is 
also necessary for ductility and energy dissipation. 

Reinforcing cover plates: HSS and pipe braces are subject to net section fracture at the gusset 
plate slots (Uriz and Mahin, 2004). This brittle failure mode can be eliminated by adding 
reinforcing cover plates to the sides of the HSS without the slots, such as the ones shown in 
Figures 8.4.1-1 and 8.4.1-2. For pipes, the reinforcing plates can be oriented at right angles to 
the pipe and appear like stiffeners. Additional information regarding the design of these plates 
can be found in Limiting Net Section Fracture in Slotted Tube Braces (Yang and Mahin, 2005). 

Cost/Disruption 

Designs that are simple and details that are not overly complicated will minimize costs 
associated with this technique. This could include maximizing the use of existing members, 
minimizing connection upgrades, and reducing the amount of field welding. 

Costs can also be reduced if disruption is minimal during construction. Installing braces at the 
perimeter frames reduces logistical issues associated with working in confined spaces and 
temporary removal of the nonstructural elements. Noise associated with this type of work is loud 
and disturbing to the tenants if the building is occupied while the work is being performed. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames 

Construction Considerations 

See Section 8.4.1 for general discussions of welding issues, removal of existing nonstructural 
and structural elements, and construction loads. Connecting members, such as gusset plates, that 
are not being replaced should be protected when braces are removed. 

Proprietary Concerns 

Braces used in buckling-restrained braced frames (BRBF) are proprietary. There are a limited 
number of manufacturers of the braces used in BRBF. 

9.5 References 
AISC, 2005, Seismic Provisions for Structural Steel Buildings, American Institute of Steel 
Construction, Chicago, IL. 

Astaneh-Asl, A., Goel, S.C., and R.D. Hanson, 1986, Earthquake-Resistant Design of Double 
Angle Bracing, Engineering Journal, Vol. 23, No. 4, American Institute of Steel Construction, 
Chicago, IL. 

FEMA, 2000a, Recommended Seismic Evaluation and Upgrade Criteria for Existing Steel 
Moment-Frame Buildings, FEMA 351, Federal Emergency Management Agency, Washington, 

D.C. 
FEMA, 2003, NEHRP Recommended Provisions for Seismic Regulations for New Buildings and 
Other Structures, FEMA 450, Federal Emergency Management Agency, Washington, D.C. 

Goel, S.C. and S. Lee, 1992, A Fracture Criterion for Concrete-Filled Tubular Braces, 
Proceedings of the 1992 ASCE Structures Congress, pp. 922-925, ASCE, Reston, VA. 

Hassan, O. and S.C. Goel, 1991, Seismic Behavior and Design of Concentrically Braced Steel 
Structures, Report No. UMCE 91-1, University of Michigan, Ann Arbor, MI. 

Horne, J., Rubbo, A., and J. Malley, 2001, AISC-LRFD Design and Optimization of Steel 
Eccentrically Braced Frames, Proceedings of 2001 Convention of the Structural Engineers 
Association of California, Sacramento, CA. 

Lee, H. and S.C. Goel, 1990, Seismic Behavior of Steel Built-up Box-Shaped Bracing Members, 
and Their Use in Strengthening Reinforced Concrete Frames, Report No. UMCE 90-7, 
University of Michigan, Ann Arbor, MI. 

Lee, S. and S.C. Goel, 1987, Seismic Behavior of Hollow and Concrete-Filled Square Tubular 
Bracing Members, Report No. UMCE 87-11, University of Michigan, Ann Arbor, MI. 

Liu, Z.,and S.C. Goel, 1988, Cyclic Load Behavior of Concrete-Filled Tubular Braces, Journal 
of Structural Division, Vol. 114, No. 7, ASCE, Reston, VA. 

Tang, X. and S.C. Goel, 1987, Seismic Analysis and Design Considerations of Braced Steel 
Structures, Report UMCE 87-4, University of Michigan, Ann Arbor, MI. 

9-13 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames 

Uriz, P. and S. Mahin, 2004, Summary of Test Results for UC Berkeley Special Concentric 
Braced Frame No. 1 (SCBF-1) Draft, Version 1.1, Department of Civil and Environmental 
Engineering, University of California, Berkeley, CA. 

Yang, F. and S. Mahin, 2005, Limiting Net Section Fractures in Slotted Tube Braces, Structural 
Steel Educational Council, Moraga, CA. 

9-14 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 10 - Type S4: Steel Frames with Concrete Shear Walls 

Chapter 10 - Building Type S4: Steel Frames with Concrete 
Shear Walls 

10.1 Description of the Model Building Type 
Building Type S4 consists of an essentially complete frame assembly of steel beams and 
columns. The floors are concrete slabs or concrete fill over metal deck. These buildings feature 
a significant number of concrete walls effectively acting as shear walls, either as vertical 
transportation cores, isolated in selected bays, and/or as a perimeter wall system. The steel 
column and beam system may act only to carry gravity loads or may have rigid connections to 
act as a moment frame to form a dual system. This building type is generally used as an alternate 
for steel moment or braced frames in similar circumstances. These buildings will usually be 
mid- or low-rise. Figure 10.1-1 shows an example of this building type. 

Figure 10.1-1: Building Type S4: Steel Frames with Concrete Shear Walls 

10-1 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 10 - Type S4: Steel Frames with Concrete Shear Walls 

10.2 Seismic Performance Characteristics 
In older buildings, the steel frame carries only gravity loads while all lateral loads are resisted by 
the concrete shear walls. In modern buildings, both lateral systems work together in proportion 
to relative rigidity. Generally, except in tall buildings, these systems tend to behave more like 
shear wall structures due to the much greater stiffness of the walls. The contribution of the steel 
moment frame to the lateral capacity of the building is a function of the number of frames and 
the detailing of the beam-column joints. See performance characteristics described in Section 

14.2 for concrete shear wall buildings and Section 8.2 for steel moment frame buildings. 
10.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
See deficiencies and techniques described in Section 13.3 and 14.3 for concrete shear wall 
buildings and Section 8.3 for steel moment frame buildings. 

10.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
See recommended techniques in Section 13.4 for concrete shear wall buildings and Section 8.4 
for steel moment frame buildings. 

10.5 References 
See references in Section 13.5 and 14.5 for concrete shear wall buildings and Section 8.5 for 
steel moment frame buildings. 

10-2 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 10 - Type S4: Steel Frames with Concrete Shear Walls 

Table 10.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S4 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce Demand Remove Selected 
Components 
Global 
Strength 
Insufficient in-
plane wall 
shear strength 
Concrete/masonry shear 
wall [8.4.2] 
Braced frame [8.4.1] 
Concrete wall 
overlay [21.4.8] 
Fiber composite wall 
overlay [13.4.1] 
Steel overlay 
Seismic 
isolation [24.3] 
Reduce flexural 
capacity 
[13.4.4] 
Insufficient 
flexural 
capacity 
Concrete/masonry shear 
wall [8.4.2] 
Braced frame [8.4.1] 
Add or enhance 
chords 
Insufficient 
frame strength 
Moment frame 
Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Strengthen beams 
[8.4.3], columns 
[8.4.3], and/or 
connections [8.4.6] 
Seismic isolation 
[24.3] 
Global 
Stiffness 
Excessive drift Concrete/masonry shear 
wall [8.4.2] 
Braced frame [8.4.1] 
Moment frame 
Strengthen beams 
[8.4.3], columns 
[8.4.3], and/or 
connections [8.4.6] 
Concrete wall 
overlay [21.4.5] 
Inadequate 
capacity of 
coupling 
beams 
Concrete/masonry shear 
wall [8.4.2] 
Braced frame [8.4.1] 
Strengthen beams 
[13.4.2] 
Improve ductility of 
beams [13.4.2] 
Remove beams 
Configuration Discontinuous 
walls 
Concrete/masonry shear 
wall [8.4.2] 
Enhance existing 
column for 
overturning loads 
Improve connection 
to diaphragm 
[13.4.3] 
Remove wall 
Soft story Concrete/masonry shear 
wall [8.4.2] 
Braced frame [8.4.1] 
Re-entrant 
corner 
Moment frame 
Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Collector [8.4.4] 
Enhance detailing 
[8.4.3], [8.4.4] 

10-3 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 10 - Type S4: Steel Frames with Concrete Shear Walls 

Table 10.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S4 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce Demand Remove Selected 
Components 
Configuration 
(continued) 
Torsional 
layout 
Add balancing walls 
[8.4.2], braced frames 
[8.4.1], or moment 
frames 
Load Path Missing 
collector 
Add collector [8.4.4] Strengthen existing 
beam [8.4.3] or slab 
Enhance splices or 
connections of 
existing beams 
[8.4.4] 
Discontinuous 
Walls 
Provide new wall support 
components to resist the 
maximum expected 
overturning moment 
Strengthen the 
existing support 
columns for the 
maximum expected 
overturning moment 
[8.4.3] 
Provide elements to 
distribute the shear 
into the diaphragm at 
the level of 
discontinuity 
[13.4.3] 
Inadequate 
shear, flexural, 
and uplift 
anchorage to 
foundation 
Embed column into a 
pedestal bonded to 
other existing 
foundation elements 
[8.4.5] 
Provide steel shear 
lugs or anchor bolts 
from base plate to 
foundation [8.4.5] 
Inadequate out-
of-plane 
anchorage at 
walls 
connected to 
diaphragm 
Tension anchors 
[16.4.4] 

10-4 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 10 - Type S4: Steel Frames with Concrete Shear Walls 

Table 10.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S4 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce Demand Remove Selected 
Components 
Component 
Detailing 
Wall 
inadequate for 
out-of-plane 
bending 
Add strongbacks [21.4.3] Concrete wall 
overlay [21.4.5] 
Wall shear 
critical 
Concrete wall 
overlay [21.4.5] 
Fiber composite wall 
overlay [13.4.1] 
Reduce flexural 
capacity of wall 
[13.4.4] 
Inadequate 
capacity of 
beams, 
columns, 
and/or 
connections 
Enhance beam-
column connections 
[8.4.6] 
Add cover plates or 
box members [8.4.3] 
Provide gusset plates 
or knee braces [9.4.1] 
Encase columns in 
concrete [8.4.2] 
Inadequate 
capacity of 
panel zone 
Provide welded 
continuity plates 
[8.4.6] 
Provide welded 
stiffener or doubler 
plates [8.4.6] 
Inadequate 
capacity of 
horizontal steel 
bracing 
Provide additional 
secondary bracing [9.4.2] 
Strengthen bracing 
elements [9.4.2] 
Reduce unbraced 
lengths [9.4.2] 
Strengthen 
connections [9.4.1] 
Diaphragms Inadequate 
in-plane 
strength and/or 
stiffness 
Collectors to distribute 
forces [8.4.4] 
Moment frame 
Braced frame [8.4.1] 
Concrete/masonry shear 
wall [8.4.2] 
Concrete topping 
slab overlay 
Strengthen chords 
[8.4.3], [8.4.4] 

10-5 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 10 - Type S4: Steel Frames with Concrete Shear Walls 

Table 10.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S4 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce Demand Remove Selected 
Components 
Inadequate 
shear transfer 
to frames 
Provide additional 
shear studs, anchors, 
or welds [22.2.7] 
Diaphragms 
(continued) 
Inadequate 
chord capacity 
Add steel members or 
reinforcement 
[8.4.3], [8.4.4] 
Excessive 
stresses at 
openings and 
irregularities 
Add reinforcement [8.4.3] 
Provide drags into 
surrounding diaphragm 
[8.4.4] 
Infill opening 
[22.2.4], [22.2.6] 
Foundations See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 

10-6 


Techniques for the Seismic Rehabilitation of Chapter 11  Types S5/S5A: Steel Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls 

Chapter 11 - Building Types S5/S5A: Steel Frames with Infill 
Masonry Shear Walls 

11.1 Description of the Model Building Type 
Building Type S5 is normally an older building that consists of an essentially complete gravity 
frame assembly of steel floor beams or trusses and steel columns. The floor consists of masonry 
flat arches, concrete slabs or metal deck and concrete fill. Exterior walls, and possibly some 
interior walls, are constructed of unreinforced masonry, tightly infilling the space between 
columns and between beams and the floor such that the infill interacts with the frame to resist 
lateral movement. Windows and doors may be present in the infill walls, but to effectively act as 
a shear resisting element, the infill masonry must be constructed tightly against the columns and 
beams. The steel gravity framing in these buildings may include truss spandrels or knee braces 
on the exterior walls, or partially restrained beam-column connections in a more extensive 
pattern. The steel frame also is often cast in concrete for fireproofing purposes. The buildings 
intended to fall into this category normally feature exposed clay brick masonry on the exterior 
and are common in commercial areas of cities with occupancies of retail stores, small offices, 
and hotels. Figure 11.1-1 shows an example of this building type. 

Figure 11.1-1: Building Type S5: Steel Frames with Infill Masonry Shear Walls 

11-1 


Techniques for the Seismic Rehabilitation of Chapter 11  Types S5/S5A: Steel Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls 

The S5A building type is similar but has floors and roof that act as flexible diaphragms such as 
wood, or untopped metal deck. This type of building will almost always date to the 1930 or 
earlier. 

Variations Within the Building Type 

The building type was identified primarily to capture the issues of interaction between 
unreinforced masonry and steel gravity framing. The archetypical building has solid clay brick 
at the exterior with one wythe of brick running continuously past the plane of the column and 
beam and two or more wythes infilled within the plane of the column and beam. The exterior 
wythe of clay brick forms the finish of the building although patterns of terra cotta, stone, or 
precast concrete may be attached to the brick or laid up within the brick. However, there can be 
many variations to this pattern depending on the number and arrangement of finished planes on 
the exterior of the building. For example, the full width of the infill wall may be located with the 
plane of the column and beam with a pilaster built out and around the column and a horizontal 
band of brick or other material covering the beam. The beam is often placed off center of the 
column, usually on the out-board side. In extreme conditions, the primary plane of the masonry 
wall may not directly engage the column at all. In these cases, strut compression must be 
transferred eccentrically through the masonry surrounding the column, reducing effectiveness. 

In some buildings the steel frame is encased in concrete, primarily for fireproofing. This 
encasement is normally reinforced with mesh and may contribute to overall frame stiffness and 
to connection strength and stiffness of partially restrained steel connections. Importantly, at the 
perimeter frames, the concrete encasement forms a smooth surface at the masonry interface and 
probably encouraged a neater fit during construction. Concrete encasement of columns also will 
assist in transferring eccentric strut loading into the column-frame system 

Hollow clay tile masonry may also be used as an exterior infill material. Although this material 
often has a very high compression strength, the net section of material available to form the 
compression strut within the frame will normally contribute a lateral strength of only a small 
percentage of the building weight. The material being brittle and the wall being highly voided, 
these walls may also lose complete compressive strength quite suddenly. Therefore, walls of 
hollow clay tile infill will probably not contribute a significant portion of required lateral 
resistance except in areas of low seismicity and/or when walls are arranged as infill on both the 
exterior and interior of the building. 

More recent buildings may have unreinforced concrete block masonry configured as an exterior 
infill wall, with a variety of finish materials attached to the outside face of the concrete block. 
Similar to hollow clay tile walls, these walls may exhibit moderate to low compressive strength 
and brittle behavior that marginalizes their usefulness as lateral elements. In addition, hollow 
concrete block exterior walls often will not be installed tight to the surrounding framing, 
eliminating infill compression strut behavior. 

11-2 


Techniques for the Seismic Rehabilitation of Chapter 11  Types S5/S5A: Steel Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls 

Floor and Roof Diaphragms 

The earliest version of this building type may include floors constructed of very shallow masonry 
arches spanning between steel beams. A relatively flat top surface is created with masonry 
rubble or light-weight cementitious fill and the floor is finished with wood sheathing. In some 
cases, the thrust from the arches is resisted by tie rods running perpendicular and through the 
steel beams. The only diaphragm action provided by such floors is the finish wood sheathing 
and the lateral flexibility of this system is incompatible with the stiff but brittle masonry arches. 

Building Type S5A will have heavy timber floors with one or more layers of sheathing forming a 
diaphragm. The flexibility of such diaphragms will often form a seismic deficiency because, 
assuming no interior shear elements, the large drift at the diaphragm mid-span will damage 
perpendicular walls and gravity framing. Specific strengthening techniques for this building type 
are not covered here. For generalized strengthening of diaphragms, see Chapter 22. 

Most typically, the floor and roof are cast-in-place concrete slabs spanning between beams. The 
concrete slab is often integral with lightly reinforced concrete surrounding each beam. This 
building type can also be found with metal deck and concrete floor slabs. 

Foundations 

There is no typical foundation for this building type. Foundations can be found of every type 
depending on the height of the building, the span of the gravity system and the site soil. The 
exterior walls are exceptionally heavy and typically will be supported by a continuous concrete 
footing or often a continuous concrete wall forming a basement space below. 

11.2 Seismic Response Characteristics 
Most steel frame infill buildings will incorporate some beam column connections with moment 
resistance, either from top and bottom chord truss connections, knee bracing, or partially 
restrained tee or angle connections. The restraint is often enhanced by cast-in-place concrete 
cover. The lateral strength and stiffness of these systems is difficult to assess, although some 
testing has been done (Roeder et al., 1996). See also Abrams (1994). Unless the perimeter infill 
is penetrated with large openings, the frame will be far more flexible than the infill. Therefore, 
both in terms of stiffness and strength, the exterior infill walls typically will form the effective 
lateral system for this building type. The effectiveness of the system depends on the size and 
extent of openings and articulation of the plane of the wall. With solid or nearly solid infill 
panels, strut action will be stiff and strong. As openings in panels increase in size, struts or 
combinations of struts cannot effectively form around the opening and the steel columns and 
beams will begin to work as a moment frame, with fixity at the beam-column joint provided by 
the masonry. For low and moderate intensity shaking, the exterior walls may provide adequate 
strength to satisfy the specified performance objective. As the shaking demand increases, the 
masonry will tend to crack and spall, losing stiffness and potentially creating a falling hazard. 
The complete steel gravity system, characteristic of this building type, is generally expected to 
provide sufficient stability to prevent collapse, particularly if designed for lateral resistance. 
However, in configurations with large height-to-width ratios, end or corner columns could fail in 
compression or at tension splices, potentially leading to partial collapse. 

11-3 


Techniques for the Seismic Rehabilitation of Chapter 11  Types S5/S5A: Steel Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

This building type is often characterized by a commercial store-front first floor with little or no 
infill at that level on one or more faces of the building. This condition can cause a soft story 
condition or a severe torsional response if open on one or two sides only. Such conditions can 
lead to concentration of seismic deformation at the open level, potentially leading to local P-delta 
failure. This open commercial story was a common feature in many buildings of this type that 
were shaken in the 1906 San Francisco earthquake, but there were no story-mechanisms 
reported. It is speculated that the soft story provided isolation for the upper stories and that the 
displacement demand, for reasons unknown, did not exceed the story capacity. In fact, there 
have been no reports of collapses or damage that suggested imminent collapse in typical U.S. 
multistory office-like steel infill buildings in strong ground motion. Earthquakes providing such 
tests include the 1906 San Francisco, the 1933 Long Beach, and to a lesser extent, the 1994 
Northridge events. In general, current seismic evaluation technology does not reach the same 
conclusion. 

11.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
See Table 11.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 
Deficiencies related to steel moment frames and masonry shear walls are shown in Table 5.3-1 
and Table 18.3-1, respectively. Selected deficiencies are further discussed below by category. 

Global Strength 

The overall strength provided by the exterior walls may be insufficient to prevent serious 
degradation and resulting amplified displacements in the building that can lead to irreparable 
damage or even instability. The strength may be limited by inadequate number of panels of 
infill, excessive openings, or masonry weak in compressive strength. The standard approach to 
such deficiencies will be to add new, relatively stiff lateral force-resisting elements such as 
concrete shear walls or steel braced frames often located on the interior between existing 
columns. Concrete walls can also be added at the perimeter on the inside face of the masonry. 
This procedure is usually conceptualized and analyzed as a concrete shear wall rather than an 
infill to the frame. 

Fiber composite layers also can be added to the face of masonry to enhance infill strut action. 
Although this technique has been tested for increasing shear strength of URM walls, little 
research is available directly on the effects of adding these layers to infill panels. 

Unless the masonry is completely doweled or connected to supporting backing, the damage state 
of the masonry wall must be estimated for the expected drifts of the combined system to 
determine if the desired performance has been achieved. 

Global Stiffness 

For this building type, the methods for adding stiffness are similar to those adding strength. 

Configuration 

Two global configurational deficiencies are common in this building type. The first is a soft and 
weak story at the street level created by commercial occupancies with exterior bays with little or 

11-4 


Techniques for the Seismic Rehabilitation of Chapter 11  Types S5/S5A: Steel Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

Table 11.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S5/S5A Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Global 
Strength 
Inadequate length 
of exterior wall 
Interior concrete walls 
[8.4.2] 
Interior steel braced 
frames [8.4.1] 
Concrete wall overlay 
[21.4.5] 
Fiber composite wall 
overlay [21.4.6] 
Excessive sized 
openings in infill 
panels 
Interior concrete walls 
[8.4.2] 
Interior steel braced 
frames [8.4.1] 
Infill selected openings 
[21.4.7] 
Concrete wall overlay 
[21.4.5] 
Fiber composite wall 
overlay [21.4.6] 
Inadequate 
columns for 
overturning forces 
Add cover plates or 
box members [8.4.3] 
Encase columns in 
concrete 
Weak or 
deteriorated 
masonry 
Interior concrete walls 
[8.4.2] 
Interior steel braced 
frames [8.4.1] 
Point outside and/or 
inside wythes of 
masonry 
Inject wall with 
cementitious grout 
Add concrete or fiber 
composite overlay on 
exterior walls pier 
and/or spandrel 
[21.4.5], [21.4.6] 
Global 
Stiffness 
See inadequate 
strength 
Configuration Soft or weak story Interior concrete walls 
[8.4.2] 
Interior steel braced 
frames [8.4.1] 

11-5 


Techniques for the Seismic Rehabilitation of Chapter 11  Types S5/S5A: Steel Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

Table 11.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S5/S5A Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Configuration 
(continued) 
Torsion from one 
or more solid 
walls 
Balance with Interior 
concrete walls 
Balance Interior steel 
braced frames 
Remove selected 
infill panels on 
solid walls 
Irregular Plan 
Shape 
Balance with interior 
concrete walls 
Balance with interior 
steel braced frames 
Load Path Out-of-plane 
failure of infill 
due to loss of 
anchorage or 
slenderness of 
infill 
Provide vertical 
strongback wall 
supports [21.4.3] 
Concrete wall overlay 
[21.4.5] 
Fiber composite wall 
overlay [21.4.6] 
Remove infill 
Inadequate 
connection of 
finish wythe to 
backing 
Add interwythe tie 
[21.4.12] 
Inadequate 
collectors 
Add steel collector on 
surface of concrete 
[12.4.3] 
Embed or add 
collector in concrete 
floor slab [12.4.3] 
Strengthen beam to 
column or beam to 
beam splices 
Component 
Detailing 
Inadequate 
columns splice for 
tension due to 
uplift force 
induced by infill 
Add splice plates 
Provide splice 
through added 
reinforced concrete 
encasement 
Inadequate beam 
column 
connection to 
resist compression 
thrust 
Strengthen 
connection in shear 

11-6 


Techniques for the Seismic Rehabilitation of Chapter 11  Types S5/S5A: Steel Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

Table 11.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S5/S5A Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Component 
Detailing 
(continued) 
Weak or 
incompletely 
filled joint 
between masonry 
and surrounding 
steel components 
Repair or fill voids 
to provide 
essentially 
continuous bearing. 
Diaphragms Flat masonry arch 
diaphragm 
Add diagonal steel 
braced diaphragm 
under floor [22.2.8] 
Remove top layers of 
floor construction and 
add concrete slab 
diaphragm 
Add tension ties to 
prevent loss of arch 
action [22.2.8] 
Foundation See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 

11-7 


Techniques for the Seismic Rehabilitation of Chapter 11  Types S5/S5A: Steel Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

no infill. This deficiency can be corrected by adding selected bays of infill or by adding shear 
walls or braced frames at this level. The second common issue is a plan torsional irregularity 
created by solid masonry walls on property lines coupled with walls with many openings on 
street fronts. If shown by analysis to be necessary, torsional response can be minimized by 
stiffening the more flexible side of the building with more infill or by the addition of lateral 
elements. In rare cases, the solid walls can be balanced with the open side by selected removal 
of panels or disengagement of the infill strut action. 

Load Path 

The primary load path issue with this building type is to assure that the mass of the exterior walls 
will not become disengaged from the frame which will both prevent infill strut action as well as 
to create a significant falling hazard on the street below. 

In-plane, the articulation of the exterior walls may result in offsets of the wall plane between 
floors. The presence of a complete load path and maintenance of confinement for strut formation 
must be reviewed in such instances. 

If new lateral load-resisting elements are added, existing slab and steel beam construction may 
need to be strengthened to provide adequate collectors. 

Component Detailing 

In order to qualify as an infill lateral force-resisting element, the infill must be installed tight to 
the surrounding steel frame. Loose or incomplete infill can be mitigated with local patching of 
the masonry or by injection of cemetitious or epoxy grout. However, unless the building is 
gutted for remodeling purposes, this procedure will be extremely disruptive. 

The detailing of the steel frame forming the confinement for the masonry is important to achieve 
infill strut behavior. The connection of beam to column must be capable of resisting the strut 
compression forces from the masonry. Many different configurations are possible, each with a 
different potential weakness, but the shear capacity of the beam-to-column connection is often 
critical. In addition, column splices may be inadequate to transfer the overturning forces created 
by strut action. Critical connections normally can be strengthened with steel plates. 

Diaphragm Deficiencies 

A wide variety of concrete diaphragms can be found in this building type. Solid slab-type floors 
will often provide an adequate diaphragm while joisted floors may include only a thin, poorly 
reinforced continuous slab with low shear capacity. The connection of slabs to exterior wall 
should be reviewed because dowels or other positive connections may not have been provided. 

See Chapter 21 on URM construction for discussion of wood diaphragms in this type of building. 

Flat masonry arch floors are problematic. The diaphragm capacity of such built up construction 
has not been established. Damage causing loss of arch action can create falling hazards or 
vertical load failures. Removal and replacement may not be feasible, either from a pure 
economic standpoint or due to historical preservation issues. The added weight of a new 

11-8 


Techniques for the Seismic Rehabilitation of Chapter 11  Types S5/S5A: Steel Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls 

concrete slab is often difficult to accommodate, even if top layers of the existing floor are 
removed. 

If space is available, a new steel diagonal frame diaphragm can be added underneath such floors. 
FRP can be layered on the masonry arches to better secure them in place. New lateral force-
resisting elements can be added to minimize the need for diaphragm action. 

Foundation Deficiencies 

No systematic deficiency in foundations should be expected solely due to the characteristics of 
this building type. 

Other Deficiencies 

Although deterioration of material, in general, is not covered in this document, it is known that 
most buildings of this type have no reliable waterproofing system for the exterior steel framing, 
particularly the columns. Significant damage to columns from water infiltration has been noted 
in several cases, and this condition should be investigated before assuming that the perimeter 
frame is a significant lateral force-resisting element. 

11.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
Most significant recommendations listed in Table 11.3-1 are similar to techniques more 
commonly associated with other building types such as steel framed buildings (S1, S2, or S4), 
unreinforced masonry bearing wall buildings (URM), or general techniques applied to concrete 
diaphragms. Details concerning these techniques can be found in other chapters. 

11.5 References 
Abrams, D.P. (Editor), 1994, Proceedings from the NCEER Workshop on Seismic Response of 
Masonry Infills, Technical Report NCEER-94-0004, National Center for Earthquake Engineering 
Research, Buffalo, NY. 

Roeder, C.W., Leon, R.T., and Preece, F.R., 1996, Expected Seismic Behavior of Older Steel 
Structures, Earthquake Spectra, EERI, Vol. 12, No. 4, Oakland, CA, pgs 805-824. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Chapter 12 - Building Type C1: Concrete Moment Frames 

12.1 Description of the Model Building Type 
These buildings consist of concrete framing, either a complete system of beams and columns or 
columns supporting slabs without gravity beams. Lateral forces are resisted by cast-in-place 
moment frames that develop stiffness through rigid connections of the column and beams. The 
lateral force-resisting frames could consist of the entire column and beam system in both 
directions, or the frames could be placed in selected bays in one or both directions. An important 
characteristic is that no significant concrete or masonry walls are present, or that they are 
adequately separated from the main structure to prevent interaction. Some buildings of this type 
have frames specifically designed for lateral loads, but also have interacting walls apparently 
unaccounted for in the design. These buildings could be classified as moment frames and the 
wall interaction would immediately be considered a seismic deficiency. Alternately, these 
buildings could be classified as Building Type C2f (Shear Wall with Gravity Frames). Older 
concrete buildings may include frame configurations that were not designed for lateral load, but 
if no walls or braces are present, the frames become the effective lateral force system and should 
be included in this building category. Buildings of this type that include integral concrete or 
masonry walls on the perimeter should be considered as Building Type C2f or C3. Floors may 
be a variety of cast-in-place or precast concrete. Buildings with concrete moment frames are 
generally used for most occupancies listed for steel moment frames, but are also used for 
multistory residential buildings. 

Figure 12.1-1: Building Type C1: Concrete Moment Frames 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Variations Within the Building Type 

The primary variation within this type is the type of frame and the number of frames included. 
Frames can range from column-girder systems of one bay on each face of the building to systems 
that employ every column coupled with two-way slabs. Frames classified by code as ductile or 
semi-ductile by code beginning in the late 1960s and early 1970s are far more constrained in 
configuration due to prescriptive rules governing girder configuration, strong column-weak 
beam, and limitations on joint shear. 

Floor and Roof Diaphragms 

The floor and roof diaphragms in this building type are essentially the same as the bearing wall 
system, and are almost always cast-in-place concrete. The diaphragms are stiff and strong in 
shear because the horizontal slab portion of the gravity system is either thick or frequently 
braced with joists. However, one way joist systems could be inadequate in shear in the direction 
parallel to the joists. Collectors are seldom in place and transfer of load from diaphragm to shear 
wall must be carefully considered. 

Foundations 

There is no typical foundation for this building category. Foundations could be found of every 
type depending on the height of the building, the span of the gravity system and the site soil. 

12.2 Seismic Response Characteristics 
This building type must be separated into older frame systems, often not even designed for 
lateral loads and including few, if any, features that would assure ductile behavior, and frames 
specifically designed to exhibit ductility under seismic loading. Rules for design of ductile 
concrete frames were developed during the 1960s. 

Older, non-ductile frame buildings, assuming an insignificant amount of concrete or masonry 
walls are present, will be far more flexible than other concrete buildings, and will probably be 
relatively weak. Most importantly, columns are often not stronger than beam or slab system, 
forcing initial yielding in these key elements. In addition, unless spiral ties were used, the 
column will typically fail in shear before a flexural hinge can form. Buildings with these 
characteristics are among the most hazardous in the U.S. inventory and are in danger of collapse 
in ground motion strong enough to initiate shear failures in the columns. Buildings of this type 
that are configured such that initial hinging occurs in the floor system will exhibit stiffness and 
strength degradation and large drifts, but unless exceptionally weak, are far less likely to 
collapse. The ratio of the inherent strength of the framedesigned for lateral loads or not 
compared to the seismic demand has a large influence on the performance, and frames in low 
and moderate seismic zones may be at less risk for this reason. 

Semi-ductile frames, with some but not all of current design features for concrete frames, likely 
will perform better, particularly if the columns are protected by basic strength and are designed 
to be flexurally controlled. However, many of these early concrete frames may be excessively 
weak and suffer from high ductility demands which could have serious consequences if a soft or 
weak story is present due to architectural configuration or column layout. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Buildings with fully ductile frames are expected to perform well, unless vertical or horizontal 
configuration irregularities concentrate inelastic deformation on certain structural components. 

12.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
See Table 12.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 
Selected deficiencies are further discussed below by category. 

Global Strength 

Although lack of ductility is the overwhelming deficiency for this building category, low 
strength may contribute to poor performance. It is difficult to add significant strength within the 
confines of the existing frames and most often new elements of braced frames or shear walls are 
added in these buildings. 

Global Stiffness 

See Global Strength. 

Configuration 

The most common configuration issue in this building type is a soft or weak story created by a 
non-typical story height. If the building is not to receive new walls or frames as part of a global 
retrofit, such configuration deficiencies can be minimized or eliminated with local strengthening 
of columns. 

Load Path 

There are no load path issues particular to this building type. 

Component Detailing 

The major deficiencies of this building type are due to inadequate component detailing, namely 
the structural components of the frame. Current requirements for ductile frames include 
capacity design techniques to assure flexural yielding in both girders and columns, as well as, for 
the most part, to limit yielding to the floor system. Retrofit procedures to obtain this ductile 
behavior of the frames are difficult, disruptive, and expensive, and are therefore seldom done. In 
high seismic zones, retrofit of these buildings is normally accomplished by adding new, stiffer 
lateral force-resisting elements that prevent significant ductility demand on the frames. 

Some research has been completed to investigate methods of retrofit for concrete moment frames 
(see Section 12.4.6), and in lower seismic zones where demands over and above gravity designs 
are not great, local strengthening and confinement of frame elements may be practical. 

Diaphragm Deficiencies 

The most common diaphragm deficiency in this building type is a lack of adequate collectors. 
The addition of effective collectors in an existing diaphragm is difficult and disruptive. Existing 
strength to deliver loads to the shear walls should be studied carefully before adding new 
collectors. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Table 12.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C1 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Global 
Strength 
Insufficient 
number of 
frames or weak 
frames 
Concrete/masonry 
shear wall [12.4.2] 
Steel braced frame 
[12.4.1] 
Concrete or steel 
moment frame 
Steel moment frame 
Increase size of 
columns and/or 
beams [12.4.5] 
Remove upper 
story or stories 
[24.2] 
Seismically 
isolate [24.3] 
Supplemental 
damping [24.4] 
Global 
Stiffness 
Insufficient 
number of 
frames or frames 
with inadequate 
stiffness 
Concrete/masonry 
shear wall [12.4.2] 
Steel braced frame 
[12.4.1] 
Concrete or steel 
moment frame 
Increase size of 
columns and/or 
beams [12.4.5] 
Fiber composite 
wrap of gravity 
columns [12.4.4] 
Concrete/steel jacket 
of gravity columns 
[12.4.5] 
Provide detailing of 
all other elements to 
accept drifts 
Supplemental 
damping [24.4] 
Remove 
components 
creating short 
columns 
Configuration Soft story or 
weak story 
Add strength or 
stiffness in story to 
match balance of 
floors 
Re-entrant 
corner 
Torsional layout 
Add floor area to 
minimize effect of 
corner 
Add balancing 
walls, braced 
frames, or moment 
frames 
Provide chords in 
diaphragm

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Table 12.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C1 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Configuration 
(continued) 
Incidental walls 
failing or causing 
torsion 
Add balancing 
walls, braced 
frames, or moment 
frames 
Uncouple incidental 
walls 
Convert incidental 
walls to lateral 
elements walls 
Remove 
incidental walls 
Load Path Inadequate 
collector 
Add or strengthen 
collector [12.4.3] 
Component 
Detailing 
Lack of Ductile 
detailing-general 
Perform selected 
improvements to 
joints [12.4.6] 
Seismic isolation 
[24.3] 
Lack of ductile 
detailing: 
Strong column-
weak beam 
Jacket columns 
[12.4.4] 
Lack of ductile 
detailing: 
Inadequate shear 
strength in 
column or beam 
Fiber composite 
wrap [12.4.4] 
Concrete/steel jacket 
[12.4.5] 
Lack of ductile 
detailing: 
Confinement for 
ductility or 
splices 
Fiber composite 
wrap [12.4.4] 
Concrete/steel jacket 
[12.4.5] 
Diaphragms Inadequate in-
plane shear 
capacity 
Concrete or 
masonry shear wall 
[12.4.2] 
Braced frame 
[12.4.1] 
Moment frame 
R/C topping slab 
overlay 
FRP overlays 
[22.2.5] 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Table 12.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C1 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Diaphragms 
(continued) 
Inadequate chord 
capacity 
New concrete or 
steel chord member 
[12.4.3] 
Excessive 
stresses at 
openings and 
irregularities 
Add chords [12.4.3] Infill openings 
[22.2.4] 
Foundations See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

12.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
12.4.1 Add Steel Braced Frame (Connected to a Concrete Diaphragm) 
Deficiencies Addressed by the Rehabilitation Technique 

Inadequate global shear capacity 

Inadequate lateral displacement (global stiffness) capacity 

Description of the Rehabilitation Technique 

Addition of steel diagonal braced frames to an existing concrete moment frame building is a 
method of adding strength and/or stiffness to the structural system. The steel braces can be added 
without a significant increase in the building weight. The new braces will commonly be some 
configuration of concentric braced frame (CBF); it is very uncommon to use an eccentrically 
braced frame (EBF) due to costs and difficult detailing issues associated with the link 
mechanism. Any of a variety of diagonal brace configurations may be used, as well as a variety 
of brace member section types. Figure 12.4.1-1 shows several common configurations. 
Common connections of the new brace to the existing concrete structure are shown in Figures 
12.4.1-2A, 12.4.1-2B, and 12.4.1-2C. 

Design Considerations 

Research basis: Design of the lateral force-resisting system for the building should account for 
the stiffness of both the braced frame system and the existing concrete moment frames. While 
basic research regarding adding braced frames at the interior of a concrete moment frame 
building has not been identified, research in the 1980s at the University of Texas at Austin on 
frames at the exterior faade demonstrated the ability of the new steel braced frames to increase 
the deformation capacity of the non-ductile concrete frames (Jones and Jirsa, 1986). A 
schematic detail of the connection used in this testing is shown in Figure 12.4.1-3. 

Braced frame  concrete frame interaction: Most designs of braced frame retrofits will be 
governed by maintaining drifts within the range of acceptability for the existing concrete 
elements. This can be accomplished by setting up a model that includes both the stiffness of the 
braced frame and of the concrete frame and meeting acceptability requirements for the 
displacements (or psuedo forces) in the concrete elements. Some engineers prefer to consider 
only the braced frames as a new lateral system, determine real drift demand for that system, and 
then check that drift for acceptability superimposed on the existing frame 

In taller buildings, the possible incompatibility between vertical cantilever behavior of discrete 
braced frames and the existing moment frames must be assessed. Existing beams or slabs, if 
unusually thick, that frame directly into the ends of new braces may restrain the global flexural 
deformation of the brace and require special consideration. Finally, due to the wide variety of 
nonlinear behavior of braced frames that is dependent on configuration and detailing, it may be 
difficult to obtain an adequate understanding of overall deformation compatibility using linear 
methods. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Figure 12.4.1-1: Typical Braced Frame Configurations 

Braced frame location: The new braces may be located on the exterior or interior of the building. 
An exterior location generally allows for easier construction access and perhaps less cost, but is 
visible, exposed to the environment and probably will impact exterior building finishes. Braces 
placed parallel to the faade can be connected to the exterior faces of perimeter spandrel beams, 
perimeter moment frames or edges of floor and roof diaphragms relatively easily, but will most 
likely cross in front of some windows. Alternatively, exterior bracing may be placed as 
buttresses, perpendicular to the existing faade. This configuration will probably require more 
extensive new collectors to deliver lateral forces from the diaphragms but may allow creation of 
new stair or elevator shafts, or perhaps additional floor area. For projects that include expansion 
of or additions to the existing building, the new braces could be located in the adjacent new 
construction, tied to the existing building. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Figure 12.4.1-2A: Typical Connection to Concrete Diaphragm 

Interior braces will most commonly be located along existing frame lines, particularly at moment 
frame bays. This will allow for best use of any existing diaphragm chords and collectors and for 
best moment frame  braced frame interaction. In some cases however, interior braces will be 
located offset from existing column-frame lines to minimize direct impact on existing structural 
or architectural components or to simplify the frame-diaphragm connections. 

The addition of new braced frames to a building will always impact the architectural character 
and functional uses of the building to some degree. Selection of preferred brace locations must 
be made considering these issues, such as space layout, corridor locations, doorways, windows, 
main M/E/P distribution runs, as well as the structural or construction considerations. 

Braced frame configuration and member section type: In most cases where diagonal steel braces 
are used to strengthen or stiffen a concrete frame building, a complete braced frame including 
horizontal beam and column members, as well as the diagonal braces themselves, is employed. 
Installation of diagonal bracing members between existing concrete columns is difficult because 
transfer of a large concentrated axial force from the concrete members through a localized 
connection with a limited number of anchors is rarely feasible. The steel columns are often 
continuous, passing through the floors, from foundations up to the roof or highest level required 
to avoid transfer of load from the steel system in and out of the concrete at each floor. In some 
cases, columns can be connected to adjacent concrete columns, but if the concrete column 
becomes part of the primary chord, reinforcing splice locations must be carefully considered. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Figure 12.4.1-2B: Typical Connection to Existing Concrete Beam 

New steel horizontal elements are similarly needed to facilitate the connections of the diagonal 
and to transfer forces from each floor into the frame. These steel elements are generally placed 
below the floor and roof diaphragms or adjacent to beams or spandrels. The diagonal steel 
braces may be placed in any of the commonly used configurations indicated in Figure 12.4.1-1; 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Figure 12.4.1-2C: Typical Connection to Existing Concrete Column 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Figure 12.4.1-3: Test Specimen Connection Detail for Braced Frame 

single diagonal or X-shaped, V-shaped, chevron (inverted-V) shaped, or super-X shaped (a 
combination of chevron and V braces in alternate stories forming a two-story X shape). Two-
story X-bracing has the advantage over V- or inverted-V-bracing should a compression brace 
buckle. In the latter configurations, the remaining tension brace has an unbalanced vertical 
component that has to be resisted by the beam. For an X-bracing configuration, even if a 
compression brace buckles, the force in the remaining tension brace is transmitted directly to the 
tension brace on the opposite side of the beam. 

Configuration will be selected based on consideration of structural issues, relative strength, 
stiffness and performance, as well as of several other issues including aesthetics, conflicts with 
doorways, corridors or windows, M/E/P systems, or the number of connections and penetrations. 
Column and beam members are often W-shapes, but may be other shapes such as channels or 
hollow structural section (HSS) tubes to improve aesthetics or to ease detailing. Diagonal 
members may be of any typically used sections including W-shapes, hollow (HSS) pipes or 
tubes, or double channels, angles or HSS tubes. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Buckling-Restrained Braced Frames (BRBFs), in which steel plates or cruciform shaped braces 
are surrounded by unbonded concrete in such a way as to prevent bucking of the brace, act 
essentially the same in tension and compression. The yield strength of a bay braced with one or 
more of these braces can be relatively accurately predicted. In situations where many, lightly 
loaded braces will be employed, sufficient global strength can be obtained by designing the 
braces to yield prior to yielding or other failure of the existing columns, preventing the need to 
retrofit the columns. 

Detailing Considerations 

Connection to existing concrete floor and roof diaphragms: A significant concern associated 
with installing a new steel braced frame in a concrete building is the connection of the beam at 
the top of the frame in each story to the underside of the existing concrete diaphragm overhead. 
The primary concern is that a relatively large shear force must be transferred from the overhead 
diaphragm into the new steel bracing below through a relatively localized connection using 
discreet anchors/bolts. The connection is generally made by one or more rows of concrete 
anchors as shown in Figure 12.4.1-2A. Typically, the anchors are threaded rods set in epoxy, but 
drilled expansion anchors may be used if they provide sufficient force transfer capacity and 
adequate testing to show they can resist cyclic loading. An alternate connection method, 
installed from the top down, consists of providing large holes in the concrete slab to expose the 
steel beam sufficiently to installed welded dowels to the top flange. The hole is then backfilled 
with cementitious or epoxy grout. In many cases however, the shear capacity of the existing 
concrete diaphragm is inadequate to deliver the relatively large shear force within the length of 
the braced frame. In those cases, a collector will be required (refer to Section 12.4.3). 

Connection to existing moment frames: New braced frames are often located on or alongside of 
the existing moment frame lines. This generally allows for better use of the existing collectors 
(beams) to deliver diaphragm forces to the bracing and, perhaps, use of the existing frame 
columns and footings to help resist overturning and uplift forces. It is generally preferable to 
locate the new braces alongside of the existing moment frames instead of as an infill within the 
width of the existing concrete frame beams and columns. 

For diagonal braces installed in an infill configuration, it is often extremely difficult to transfer 
large seismic forces from the surrounding concrete members through very localized connections 
with a limited number of discreet anchors. Also, if steel columns or vertical members are used in 
the infill frame, it is virtually impossible to provide vertical continuity from floor to floor 
through the existing concrete beams. Furthermore, if connections of sufficient strength can be 
made, the anchors must be threaded into or through the relatively densely reinforced beams and 
columns and, where collector strengthening is required, the added collector components will not 
connect directly to the new braces. In addition, physical installation and fit-up of the new braces 
and their connecting gusset plates often becomes significantly more difficult. 

These detailing difficulties can be reduced or avoided by placing the new braced frame alongside 
of the concrete moment frame. In most cases, placing the new bracing alongside an exterior 
frame will allow the greatest ease in detailing. Bracing in this location will almost always require 
installation of a complete new steel braced frame instead of only the new diagonal braces 
themselves. In this configuration, the connection of the concrete diaphragm to steel braced frame 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

can be made as discussed above or by installing anchors into the side of the adjacent concrete 
frame beam as shown in Figure 12.4.1-2B. Braced frame overturning forces are carried directly 
by the new steel column members. However, if concrete beam framing occurs in two directions, 
the new steel columns will generally need to be offset from the existing concrete column on a 
45-degree diagonal to provide continuity of the new column through the floors without 
interference with the existing concrete beams. Overturning resistance can be obtained by 
connecting the new steel column to the existing concrete column (see Figure 12.4.1-2C1) and 
footing. For cases where the new braced frame can be placed on the exterior of the building, the 
new steel columns can be continuous and the connections to the adjacent concrete columns or 
pilasters can be made with relative ease. If diaphragm collector strengthening is required, the 
additional collector can be installed alongside of the existing beam line and can be connected 
directly to the new braced frame. 

Exterior bracing at offset columns: There are some buildings where the exterior columns 
protrude farther out than the exterior beams. The Jones and Jirsa (1986) research can be applied 
in these situations, where the new steel framing is placed adjacent to the beams and in the plane 
of the outer portion of the protruding concrete column as shown in Figures 12.4.1-3 through 
12.4.1-5. The primary challenges lie in connecting the two types of frames and delivering loads 
into the braced frames. As an alternative to drilling numerous holes for bolts or dowels into the 
concrete columns, steel lugs can be provided at each floor. In this approach, steel pipes are 
inserted through cores drilled through the concrete columns and filled with grout. Next, the 
pipes are welded to steel plates on the sides of the concrete columns, which then provide surfaces 
for welding to the columns of the braced frames. An example of this connection is shown in 
Figure 12.4.1-4. If required, horizontal forces can also be transferred directly to the braced 
frames through the braced frame beams. The beams are welded to steel plates, which are 
connected to the concrete slab or beams at the building perimeter with dowels, bolts, or lugs, as 
shown in Figure 12.4.1-5. 

Exposed exterior braced frames require simple and clean connections that fit the architectural 
character of the building. Use of W-shapes for the braced frame members can eliminate gusset 
plates and allow direct connection of the members through complete joint penetration welds. 
Shop welding of the connections and on-site prefabrication of the braced frames will minimize 
field welding on the structure. W-shapes also simplify other architectural issues by not allowing 
rainwater or debris to accumulate. 

Installation of additional collectors: Installation of new braced frames in a concrete frame 
structure, especially in one with a distributed frame system, will often result in increased 
diaphragm demands at the individual braces. An advantage of locating the new brace at an 
existing frame line is that the existing beams can then be used as a collector. However, 
insufficient continuity and/or laps of reinforcing steel combined with highly concentrated 
diaphragm demand may still require strengthening of the existing collector (refer to 
Section 12.4.3). 

Footings: Addition of steel braced frames to an existing building will almost always require 
construction of new footings, or augmentation of existing ones, to resist the concentrated 
overturning demand. In many cases, the overturning uplift demand will require installation of tie 

12-14 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Figure 12.4.1-4: Braced Frame to Concrete Column Connection 

downs. Alternatively, the new frame can be located between two existing column frame lines, 
instead of directly on or along one frame line, and new foundations or grade beams can be used 
to engage more than one existing column to resist the uplift demand. 

Cost/Disruption Considerations 

The cost and level of disruption associated with installation of steel bracing is generally less than 
that of shear walls. The number of penetrations that need to be cut through the existing concrete 
structure and of drilled dowels and anchors may be less than for the shear wall alternative, and 
the work is generally not as wet or messy. Also, it will not be necessary to prepare any existing 
concrete surfaces that will be in contact with new steel members. The new members are discrete 
and welded or bolted connections are localized. However, there will be noise and vibrations 
resulting from the required cutting and drilling that will make continued occupancy difficult. 

12-15 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Figure 12.4.1-5: Braced Frame to Slab Connection 

Invariably, some architectural and M/E/P system components will require relocation or 
replacement. 

Construction Considerations 

The primary construction consideration will be fit up and installation of the steel braces and their 
connections. The desire to limit the number of splice connections must be balanced against the 
difficulties of installing longer members such as multistory columns. Installation of the diagonal 
braces will require careful planning and will often require member splices in the field. 
Installation of drilled threaded rod or expansion anchors will require some precision and 
extensive use of templates and oversized holes, to assure proper fit with the steel members. In 
some cases, the steel members themselves could be used as the template for the anchors. 

Proprietary Concerns 

In general, there are no proprietary concerns related to installation of steel diagonal braced 
frames in a building. The one exception, however, occurs if buckling-restrained braces are used. 

12-16 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

12.4.2 Add Concrete or Masonry Shear Wall (Connected to a Concrete 
Diaphragm) 
Deficiencies Addressed by the Rehabilitation Technique 

Inadequate global shear capacity 

Inadequate lateral displacement (global stiffness) capacity 

Description of the Rehabilitation Technique 

Addition of shear walls to an existing concrete frame building is a common method of adding 
significant strength and/or stiffness to the structure. The new walls may be of cast-in-place 
concrete, shotcrete or fully grouted concrete masonry unit (CMU) construction. 

Design Considerations 

Research basis: No research focused on the overall effects of adding shear walls to existing 
concrete frames has been identified. The effects of surface preparation, concrete strength, and 
interface reinforcement on interface shear capacity between new and existing concrete were 
examined by Bass, Carrasquillo, and Jirsa (1985). These tests indicate that surfaces prepared 
with heavy sandblasting exhibit shear capacities greater than or equal to those exhibited by 
chipped surfaces or surfaces prepared with shear keys. Increased concrete strength resulted in 
increased interface shear capacity for chipped surfaces and those prepared with shear keys, but it 
had little effect on the shear capacity of interface surfaces prepared by sandblasting. Specimens 
in which drypack mortar was used exhibited a significantly smaller shear capacity than those 
where new concrete was cast directly against the interface. Increasing the amount or embedment 
depth of reinforcement across the interface resulted in greater interface shear capacity. 

Frame-wall interaction: Most designs of shear wall retrofits will be governed by maintaining 
drifts within the range of acceptability for the existing concrete frame elements. This can be 
accomplished by setting up a model that includes both the stiffness of the shear walls and of the 
concrete frame and meeting acceptability requirements for the displacements (or psuedo forces) 
in the concrete frame elements. Some engineers prefer to consider only the shear walls as a new 
lateral system, determine the expected drift demand for that system and then check that drift 
superimposed on the existing frame for acceptability. 

In taller buildings, the possible incompatibility between vertical cantilever behavior of discrete 
shear walls and the existing concrete frames must be assessed. Existing beams or slabs, if 
unusually thick, that frame directly into the ends of new walls may restrain the global flexural 
deformation of the wall and require special consideration. Two shear walls are often purposely 
placed in line and connected by a short beam to form a coupled shear wall system. In this 
system, the coupling beams are specially designed to accept significant inelastic deformations. 
Seldom can two such walls utilize existing beams as coupling beams due to inadequate detailing. 
Thus coupling beams, when employed, are installed new, as part of the system. 

Frame-wall configuration: A primary design consideration is determination of whether or not the 
existing concrete frames may be used as an effective part of a combined system. Are the existing 
frame columns strong enough and/or well detailed enough to serve as the chord/boundary 
member of a shear wall without improvement? Are the frame beams detailed well enough to 

12-17 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

serve as coupling beams or to be incorporated into the wall itself? These considerations may 
limit the choices of wall-frame physical relationship: that is, should the walls be placed 1) 
within the plane of the existing concrete frames, 2) as vertically continuous walls alongside of, 
and joined to, the existing frames, or 3) as separate vertical elements independent of the frames? 
The first alternative is often best avoided as noted in the Detailing Considerations discussion 
below. Considering alternates 2 and 3, it must be determined if the frames are capable of 
becoming part of the shear wall (primarily as chord elements) or if it is beneficial to prevent 
direct interaction by placing the shear walls free of the existing concrete frame elements. In 
some cases, it is not feasible to stiffen the building into the range of acceptable deformation of 
the existing frames, and improvement in deformation capacity may be required in addition to the 
addition of new walls. 

Wall location: The new walls may be placed on the exterior or interior of the building. An 
exterior location generally allows for easier construction access and perhaps less cost, but is 
visible, exposed to the environment and may impact exterior building finishes. Walls placed 
parallel to the faade can be connected to the exterior edges of floor and roof diaphragms or 
perimeter concrete frames relatively easily, but will most likely require closure or reduction in 
size of some windows. Alternatively, exterior walls may be placed as buttresses perpendicular to 
the existing faade. This configuration will probably require more extensive new collectors to 
deliver lateral forces from the diaphragms but may allow creation of new stair or elevator shafts, 
or even of additional floor area. For projects that include expansion of or additions to the existing 
building, the new walls could be located in the adjacent new construction. 

Interior walls located along frame lines, particularly at moment frame bays will often allow for 
best use of any existing diaphragm chords and collectors. Beams that frame directly into the ends 
of new walls may behave like coupling beams as described above. In some cases, interior walls 
are better located offset from existing column-frame lines to minimize direct impact on existing 
structural or architectural components or to simplify the wall-diaphragm connections. 

The addition of shear walls to a building will always impact the architectural character and 
functional uses of the building. Selection of preferred wall locations must be made considering 
these issues, such as space layout, corridor locations, doorways, windows, main M/E/P 
distribution runs, as well as the structural or construction considerations. 

Detailing Considerations 

Connection to existing concrete floor and roof diaphragms: Arguably, the most significant detail 
associated with installing a new shear wall in a concrete building is the connection at the top of 
the new wall to the underside of the existing concrete diaphragm overhead. The construction 
joint must be made tight, without any gapping, to facilitate transfer of shear forces from the 
overhead diaphragm into the new wall below and to minimize the possibility of joint slip. See 
the discussion under the Research basis section. 

Typical details of this connection for a new cast-in-place concrete wall below an existing 
concrete flat slab are shown in Figures 12.4.2-1A and 12.4.2-1B. The vertical dowels must be 
sufficient to transfer forces from the existing diaphragm and from the new wall above (if it 
exists), to the lower wall. Shears can also be transferred across this joint with large diameter 

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Figure 12.4.2-1A: Concrete Wall Connection to Concrete Slab 

pipes or structural shapes. The holes made through the existing slab must serve not only to 
install the dowels, but also to allow for placement and consolidation of the wall concrete. The 
concrete head created by placement up to the top of slab coupled with cleaning and roughening 
the existing concrete contact surface by either sandblasting or chipping will provide the best joint 
available. The larger holes through the slab will also be more like intermittent shear keys. The 
holes should be drilled or made with impact tools instead of saws or core drills to avoid cutting 
or damaging existing slab reinforcement. Prior to cutting the holes, temporary shores may be 
required below the slab along each side of the row of holes. The concrete should be placed 
through the slab openings into the forms below, up to top of slab, to provide some head on the 
joint at the underside of the diaphragm. 

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Figure 12.4.2-1B: Concrete Wall Connection to Concrete Slab  Partial Elevation View 

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If the new wall is shotcrete, special care is required by the nozzle operator when placing the 
shotcrete directly at the underside of the slab to provide a tight, well-bonded joint free of 
rebound or gaps. To minimize the possibility of creating a gap due to sagging, the last lift of 
shotcrete should be a short one. In the end, however, such a well-bonded joint is often not 
achieved at the slab soffit and remedial work, similar to crack-injection repairs, is likely to be 
needed. As an alternative, the holes through the slab needed for placement of the vertical rebar 
dowels could be made oversized, sufficient to allow placement of pourable, cementitious grout at 
the top of the shotcrete wall below, similar to the cast-in-place concrete alternative. 

For CMU wall construction, the masonry units will typically be constructed up to within one or 
two courses of the overhead slab soffit, leaving enough of a gap to allow placement of the upper 
lift of grout. Preferably, the gap should be formed and grouted from above through holes in the 
slab similar to cast-in-place concrete alternative described above. Consolidation of the upper lift 
of grout should be performed through holes in the slab above. Although the gap can be dry 
packed from below, this is a considerably less effective alternative as confirmed by the research 
results noted above. 

Regardless of whether the new wall is cast-in-place concrete, shotcrete or CMU, some shrinkage 
or sagging will probably occur creating a crack at the joint. To account for the resulting 
reduction in effective aggregate interlock along this joint, it may be prudent to use a lower 
coefficient of friction, and increase the size the vertical dowels. 

Figure 12.4.2-1C shows the conditions where the existing concrete diaphragm is in a pan joist or 
waffle slab system instead of a flat slab. For these types of floor or roof systems, the joists or 
waffle ribs must be preserved to avoid shoring. However, there is likely to be more flexibility in 
the extent of the openings that can be made through the slab between the joists/ribs, and 
temporary shoring will generally not be required. Where the new wall is parallel to the joists, it is 
preferable to locate the wall offset from the joist as shown in the detail. The holes in the slab may 
be made as intermittent keys, similar to the flat slab condition discussed above, or they can be 
made as relatively long slots or as a continuous opening the length of the wall. Additional 
diaphragm to wall shear transfer capacity can be obtained by doweling into the side of the 
adjacent rib. 

Where the new wall is perpendicular to the joists, or at a waffle slab condition, the slab can be 
removed between the ribs as indicated in Figure 12.4.2-1D. Since installation of continuous 
horizontal wall bars through the perpendicular ribs is generally not possible, installation of one 
or two horizontal hoop ties may be required at the upper portions of the wall between the ribs. 
For CMU wall construction, the masonry will stop below the joists or ribs, and the large vertical 
gap up to the slab, between the ribs, will be completed with poured concrete. 

Connection to existing frames: New walls are often located on or alongside existing frame lines. 
This generally allows for better use of the existing diaphragm collectors (beams) and of the 
existing frame columns as wall chords or boundary elements. It is almost always preferable to 
locate the wall alongside the frame beams instead of as an infill, within the width of the frame 

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Figure 12.4.2-1C: Connection of Concrete Wall to Concrete Joists or Waffle Slab 

beams and columns. When placed alongside the frame, the wall-diaphragm connections are as 
discussed above, and additional shear transfer and wall chord capacity can be obtained by 
doweling into the side of the beam and the column, respectively. Also, if diaphragm collector 
strengthening is required, the additional collector can be installed alongside the existing beam 
line and will be lead directly to the new wall. In the infill configuration, the vertical wall 
dowels must be threaded through the relatively densely reinforced beams, concrete placement 
and consolidation becomes significantly more difficult, and additional collectors do not connect 
directly to the new wall. 

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Figure 12.4.2-1D: Concrete Wall Connection to Waffle Slab  Partial Elevation View 

If the existing columns have sufficient strength and appropriate reinforcement detailing, they 
may be used as the wall chord or boundary element, by doweling into the column. The 
effectiveness of this is limited by the amount of doweling that can be installed. In many cases, 
however, the existing column will require strengthening or jacketing, or new wall chords will be 
needed. 

Installation of additional collectors: Installation of shear walls in a frame structure, especially in 
one with a complete frame system, will result in increased diaphragm demands at the individual 
walls. An advantage of locating the new wall at an existing frame line is that the existing beams 
can then be used as a collector. However, insufficient rebar continuity and/or laps combined with 
highly concentrated diaphragm demand may still require strengthening. Refer to Section 12.4.3. 

Footings: Addition of concrete or masonry shear walls will almost always require construction 
of new footings, or augmentation of existing ones, to support the added weight as well as to 

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resist the increased and/or concentrated overturning demand. In many cases, the overturning 
uplift demand will require installation of tie downs. Where the new wall is located between 
column frame lines, instead of directly on or along one frame line, new foundations can be used 
to engage more than one column to resist the uplift demand. 

Cost/Disruption Considerations 

In general, shotcrete walls are less expensive than cast-in-place concrete because at least one side 
of the wall forming is eliminated. If shotcrete can be applied against an existing wall at stair, 
elevator or mechanical shafts, the cost savings of shotcrete is even greater. CMU walls are 
generally less costly, per square foot, than either shotcrete or cast-in-place concrete walls. 
However, CMU walls may not provide comparable strength or stiffness, requiring the addition of 
more linear feet of CMU walls than either cast-in-place concrete or shotcrete walls. 

Construction of new shear walls in an existing building can be very disruptive to any building 
occupants. Noise, vibration, and dust associated with many operations, especially cutting holes 
through and drilling dowels into concrete, can be transmitted throughout a concrete structure. 
Placing cast-in-place concrete, shotcrete or even grouted masonry is a wet process and very 
messy. Shotcreting in an enclosed area creates differential pressures that can spread debris 
beyond nominal construction barriers. Also, excavation and drilling operations and the use of 
mechanized and/or truck mounted equipment associated with installation of new foundations can 
be very disruptive. 

Construction Considerations 

The existing concrete surfaces to be in contact with the new concrete walls should be cleaned of 
all finishes, paint, dirt, or other substances and then be roughened to at least attempt to provide 
1/4 minimum amplitude aggregate interlock at joints and bonded surfaces. At overhead joints 
where such preparations may be less effective, as discussed in the Detailing Considerations 
section above, additional dowels can be used with less roughening. 

For shotcrete applications, separate trial test panels at the overhead joints should be included 
with the normal preconstruction test panels. These test joints should be cored to inspect the 
adequacy of the surface preparation and the joint bond. Nozzle operators should have several 
years experience with similar structural seismic improvement applications. 

In addition to the usual concrete/shotcrete core sampling and testing, the overhead joints should 
be cored to allow inspection of the joint quality and determine whether or not repairs are needed. 

For CMU shear walls, practical limitations on placement of wall reinforcing steel must be 
considered. In particular, use of seismic comb type of joint reinforcement (a prefabricated 
mesh of welded wire reinforcement used as transverse reinforcement at boundary elements of 
CMU walls) has often proven to be very difficult to install and the resulting rebar congestion 
interferes with grouting operations. 

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12.4.3 Provide Collector in a Concrete Diaphragm 
Deficiencies Addressed by the Rehabilitation Technique 

Inadequate or missing collector 

Inadequate diaphragm chord capacity 

Description of the Rehabilitation Technique 

Addition of a new collector or strengthening of an existing collector is often needed when new 
steel braced frames or concrete shear walls are added to an existing building. The new collector 
must extend as far as necessary, often one or more bays from one or both ends of the new brace 
or wall, to draw the required shear demand from the existing diaphragm. The new collector will 
be constructed of reinforced concrete or steel, generally depending on whether the general 
building upgrade involves installation of new concrete shear walls or steel braced frames. The 
new collector will most often be installed at the underside of floor. At roofs, the collector may be 
placed either from below or above the roof. 

In cases where the existing diaphragm chord is absent or inadequate, the mitigation approach 
will be similar to that used for collectors. 

Design Considerations 

Research basis: For new reinforced concrete collectors, see the discussion of tests by Bass, 
Carrasquillo, and Jirsa (1985) in Section 12.4.2. 

For new steel collectors, Jimnez-Pacheco and Kreger (1993) tested single anchor connections 
between existing concrete and new steel members in order to examine shear transfer along the 
interface between these two elements. Results indicate that sandblasting the steel surface and 
applying a layer of epoxy at the interface between steel and concrete can substantially increase the 
force level at which the interface begins to slip. Also, the use of spring washers may reduce long-
term anchor bolt relaxation, maintaining the first-slip force capacity over time. For applications 
where significant inelastic deformations are expected, a thick layer of nonshrink grout between the 
steel and concrete was found to increase deformation capacity, though it decreased ultimate strength 
slightly. Filling the annulus between the bolt and washer with epoxy resulted in greater connection 
stiffness than that exhibited by specimens with unfilled annuli or those filled with nonshrink grout. 

Material selection - reinforced concrete or steel: In reinforced concrete buildings with some sort 
of concrete slab floor system, especially one with joists, waffle ribs or beams crossing the path of 
the collector, the most common material choice for the new collector is reinforced concrete. 
Often, this choice is made because concrete is aesthetically compatible with the surrounding 
structure, especially in a condition exposed to view. However, concrete is selected principally 
because it is compatible with the deformation characteristics of the diaphragm it is connected to. 
A concrete collector is bonded to, and is integral with, the concrete slab diaphragm and the strain 
deformations of the collector are the same as the deformations of the diaphragm system. At a 
steel plate collector, the elongation of the plate is not compatible with the diaphragm slab. As the 
collector load accumulates towards the connection to the new wall or brace, the elongation of the 
plate accumulates as well. The threaded rod anchors connecting the plate collector to the 

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diaphragm in the zone of greatest elongation can become overloaded to failure by the plate 
bearing on the bolts. This can lead to a zipper-like failure mode as the adjacent anchors assume 
the load of the failed anchors and become overloaded in turn. This behavior can occur even at 
relatively short collectors if the elongation exceeds the available annulus gap around the anchor. 
To avoid this, special detailing is required as discussed in the Detailing Considerations section 
below. 

Impact on architectural and M/E/P systems and components: A new collector often must extend 
one or more entire bays away from the new wall or brace in order to draw the necessary load 
from the existing concrete diaphragm. Installation of the new collector at the underside of the 
existing floor slab impacts any existing ceilings, partitions, ductwork, plumbing, lighting, etc., 
located along its entire length. As a result, the new collectors will often have a greater impact on 
the buildings other systems than the new walls or braces themselves. Furthermore, consideration 
of these impacts will often affect placement of the new walls or braces. In many cases, the new 
walls and their associated collectors are located along the exterior edge of the building 
specifically to avoid or minimize these impacts on other building systems, especially in a case 
where building occupancy is maintained during the construction. 

In some cases, it may be possible to locate the new collector at the top surface of the existing 
diaphragm. At roofs, new collectors can be placed on top of the roof diaphragm, provided that 
any conflicts with roof mounted equipment, pads or penthouses can be accommodated or 
avoided. More importantly, placement of collectors on top of the roof slab requires careful 
consideration of the impact on roof drainage and waterproofing systems. At floors, the 
opportunity exists if a new concrete topping or structural overlay is proposed. In this case, the 
reinforcement for the collector can be embedded in the topping. Also, if a new raised floor 
system is being installed, it may be possible to locate a new collector in the space beneath the 
new floor. 

Weight of new collector: The gravity load capacity of the existing slab, waffle ribs, joists or 
adjacent beams must be adequate to support the additional weight of the new collector, 
especially for a new concrete collector that may represent a considerable load. In some cases, 
the new collector may need to be designed to support itself as it spans between existing girders 
or columns. In others, it may be required to adjust the location of the collector, and the new shear 
walls, if the existing floor or roof slab cannot support the new loads. 

Detailing Considerations 

Connection of a reinforced concrete collector to existing concrete diaphragms or collectors: A 
typical detail of the installation of a new reinforced concrete collector to the underside of an 
existing concrete slab diaphragm is shown in Figure 12.4.3-1. The primary considerations are to 
provide a good bond between the new and existing concrete and to provide adequate access ports 
for concrete placement and consolidation. The contact surface must be thoroughly cleaned and 
roughened for good shear transfer performance. It is best to place concrete from above through 
pour ports made in the diaphragm. The ports will need to be at least 4 inches in diameter. Care 
must be taken to locate existing diaphragm reinforcement before cutting the ports to avoid 
cutting any bars in what is likely to be a lightly reinforced slab. 

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Figure 12.4.3-1: Concrete Collector at Concrete Slab 

The required length of collector will be determined primarily by the existing diaphragm shear 
capacity. Lightly reinforced diaphragms can deliver only a limited load per foot, requiring long 
collectors. Also, for thin diaphragm slabs, the shear capacity of each drilled dowel will be 
limited, requiring more dowels. If the collector crosses any existing beam or girder, a splice must 
be made through the existing member. Horizontal holes can be drilled through the member and 
dowels installed to lap with the main collector reinforcing bars on each side. Care must be taken 
to avoid cutting any reinforcement, either main longitudinal bars or stirrups, in the existing 
beam. 

If the existing floor or roof diaphragm is a waffle or pan joist system, the continuous collector 
will almost always be placed below the ribs, as shown in Figure 12.4.3-2, to avoid excessive 
drilling and rebar splicing. In this condition, the voids between the ribs, above the dropped 
collector, will be filled with reinforced concrete. Advantages of this condition are that the drilled 
dowels can be installed into the sides of the ribs instead of the relatively thin cover slab, and 
making pour ports through the slab is likely to be less problematic. Also, although the new 
collector may weigh more in this condition, the waffle or joist ribs are much more likely to have 
adequate strength to support the added weight. 

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Figure 12.4.3-2: Concrete Collector at Waffle Slab 

In many cases, the new collector will occur along an existing beam or girder line. Often, this will 
occur if the task is to strengthen a diaphragm edge chord or if the new walls occur at the 
buildings exterior. Figure 12.4.3-3 shows two generic conditions that can be used in this case. In 
this condition, the dowels will always be placed into the beam, and the combined beam-collector 
member will easily be designed support the added weight. However, special care should be taken 
to avoid cutting any beam stirrups or slab diaphragm reinforcement, especially at an exterior 
edge condition, with the pour holes. 

In any of these collector configurations, a significant portion of the main reinforcement can be 
provided by a steel plate instead of by bar reinforcement (refer to Figure 12.4.3-2). This option 
may be best for conditions of very high loads, where installation of a high strength steel plate 
may be preferred over placement of many large bars. 

Connection of a steel collector to existing concrete diaphragms: Steel plate also may be used as 
the collector in lieu of a reinforced concrete member. A steel collector will have to be installed in 

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Figure 12.4.3-3: Concrete Collector at Existing Beam 

manageable sections, generally about 10 to 20 feet in length, and will be connected to the 
concrete diaphragm with drilled threaded rod anchors set in adhesive or epoxy. In almost all 
cases, the steel plates will be installed at the top of the diaphragm as shown in Figure 12.4.3-4. 
Although possible, it is extremely difficult to install heavy plate sections, connect the bolts and 
make the necessary welded splices from below. 

As discussed in the Design Considerations section above, the primary concern with a steel plate 
collector is its lack of strain compatibility with the concrete diaphragm, unless the collector is 

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Figure 12.4.3-4: Steel Plate Collector 

very short. The strain deformation of a steel collector will vary from zero at its free end to a 
maximum at the connection to the wall or brace while the concrete diaphragm will not 
experience similar deformations. In effect, the steel collector will stretch like a very stiff rubber 
band relative to the concrete diaphragm. This relative deformation is difficult to accommodate, 
especially in relatively long collectors. To do that, several conditions must be considered. First, 
the various plate sections of the collector must be stepped in size so the strain is distributed 
relatively equally along the length of the collector. Second the plates must be sized to limit the 
maximum elongation to a reasonable amount of about one or two inches. Third, the threaded rod 
anchors must be installed in slotted holes to allow the design elongation to occur without bearing 
on and overloading the anchors. Fourth, to allow the slip to occur between the collector and 
diaphragm, load transfer must be accomplished by friction using specially calibrated spring 
washers to generate the appropriate clamping force in the anchors. 

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Cost/Disruption Considerations 

Collectors have significant cost/disruption impact in a retrofit project primarily due to their 
length. They impact many building systems over a relatively large area compared to the impact 
associated with the walls themselves. This is especially true if general renovation of architectural 
and M/E/P systems is not included in the project. Thus, any available means of reducing 
collector length will probably be cost effective. A collector installed at the exterior edge of a 
diaphragm will generally be less costly than one installed in the interior and one installed above 
the diaphragm will be easier to install and, generally, less costly than one installed from below. 
However, installation of any collector can be very disruptive to any building occupants, due to 
the noise and vibration caused by drilling and coring through concrete, as well as the likely need 
to relocate various utilities and service distribution systems. 

A comparison of the cost between reinforced concrete and steel plate collectors is very difficult 
to make. In general, the cost of either type of collector installed from below the diaphragm will 
likely be similar, because so much of the cost will be related to the impact on other systems. The 
cost of a steel plate collector may be less than one of reinforced concrete, but only for collectors 
installed from above the diaphragm, and particularly on a roof. 

Construction Considerations 

Existing concrete surfaces to be in contact with new concrete or steel plate collector should be 
cleaned of all finishes, paint, or other substances that could impair bond and shear transfer 
capability. Surfaces to receive new concrete should be roughened to provide  amplitude 
aggregate interlock to prevent slip. However, since slip is expected to occur as a steel collector 
elongates, only light sandblasting may be required to assure development of the appropriate 
friction. 

Installation of grouted anchors and/or dowels for steel plate collectors will require relative 
precision. They must be installed at the middle of the long slotted holes, with only a small 
tolerance, to allow the plate to elongate without bearing on the anchor. If existing rebar is 
encountered at a location, that location should be abandoned and the anchor installed in an 
available adjacent or nearby slot. Since it is reasonable to expect that this will occur with some 
frequency, a substantial amount of extra slotted holes must be available. For instance, if anchors 
are required at 12 on center, slotted holes should be provided at 6 on center. 

The complete penetration welded splices of the relatively thick steel plate collectors are likely to 
be problematic. Making one-sided complete penetration welds in relatively thick plates will 
cause the plates to curl. To control this, the plate sections may need to be anchored down, with 
anchors placed in addition to the required shear anchors, and welded in place. Also, removal of 
backup bars will be difficult or impossible. Notches may be made into the concrete slab, and any 
remaining gap between the bottom of the steel plate collectors and the concrete diaphragm slab 
must be filled with grout to assure adequate friction at the concrete-steel interface. 

Overhead construction of reinforced concrete collectors will require careful consideration and 
planning of how the reinforcing steel is placed and secured, prior to closing up the forms from 
below. Making the pour access ports and any sleeve holes for continuous rebar will require 

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careful scanning of the slabs, waffle ribs, and beams to locate existing reinforcement to avoid 
cutting any existing rebar. 

While the inspection, sampling, and testing required for reinforced concrete collectors is not 
particularly different from what is required for other seismic force-resisting reinforced concrete 
work, some special considerations do occur for steel plate collectors. The welded splices will 
require careful, non-destructive testing and thorough inspection. The shear anchors must be 
located at the middle of the slotted holes with some precision, and they must be extensively 
proof tested in tension. The holes with anchors must be free of any grout that could reduce the 
range of slip. Any gap between the bottom of the plate and the concrete slab must be grouted. 
The installation of the spring washers must be carefully inspected and tested to assure 
development of the design clamping force. 

Proprietary Concerns 

The basic materials are generic. 

12.4.4 Enhance Column with Fiber-Reinforced Polymer Composite Overlay 
Deficiencies Addressed by the Rehabilitation Technique 

Inadequate shear capacity 

Inadequate concrete compression strain and stress capacity due to lack of concrete 

confinement 

Inadequate lap splice 

Description of the Rehabilitation Technique 

The use of a fiber-reinforced polymer (FRP) overlay with columns has proven to be an efficient 
rehabilitation technique in both the building and bridge construction industries. Columns are 
overlaid with unidirectional fibers in a horizontal orientation, thus providing shear strengthening 
and confinement similar to that provided by hoops and spirals used with circular columns, and 
stirrups and ties used with rectangular columns. The confinement enhances the concrete 
compression characteristics, provides a clamping action to improve lap splice connections, and 
provides lateral support for column longitudinal bars. 

The preferred strength hierarchy for a building type structure is strong-column, weak-beam. 
Where the strength hierarchy results in weak-column, strong-beam (and is not considered 
acceptable by the designer due to, perhaps, concern for a soft story mechanism), the use of FRP 
overlay as flexural strengthening should not be used, unless there are extenuating circumstances 
and a very detailed analysis and design are performed. The uncertainty of strain compatibility 
between the FRP and column longitudinal bars and between the FRP and substrate, the lack of 
vertical strain capacity as a result of using FRP as longitudinal reinforcement, and the anchorage 
of the FRP at column ends and at points of contra-flexure deem this approach as undesirable. 
Other techniques presented in this document should be used in this situation. 

See Section 13.4.1, Enhance Shear Wall with Fiber-Reinforced Polymer Composite Overlay, 
Fiber-Reinforced Polymer Composite Overview, for background information. 

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Design Considerations 

Research basis: Seible and Innamorato (1995) is one of the original ground-breaking research 
efforts of this topic and serves as an excellent source for understanding and design equations. 

The primary deficiencies of a column are typically the lack of shear strength capacity and post-
yield shear deformation capacity, as observed during many earthquake events and in laboratory 
testing. For shear assessment, two column locations need to be evaluated for shear: the end 
region within the plastic hinge zone (where the concrete shear strength degrades), and the region 
away from the flexural hinges, where there is no concrete shear strength degradation. 

The FRP overlay provides confinement to enhance the concrete strain and stress capacity. 
Confinement is more effective for circular sections than rectangular sections. For circular 
sections, the passive radial pressure exerted by the FRP overlay on the gross concrete section, 
which is a result of the concrete lateral dilation, provides confinement. Dilation (similar to the 
concrete splitting action when performing a pure axial compression test on a concrete cylinder) 
occurs as result of the compression force, which is influenced by the level of axial load and 
flexural forces. For a rectangular section, dilation is only arrested at the corners of the section, 
thus relying on the concrete to arch between the corners, resulting in a reduced concrete core 
size. Due to the lack of effective confinement by the FRP, it is recommended to limit the 
rectangular section to a 1.5 depth-to-width aspect ratio and a width or depth dimension less than 
36 inches, unless a special study is performed. 

The confinement afforded by this technique does marginally increase the flexural strength and 
stiffness of the column, but not to the degree of concrete jacketing. The marginal increase is due 
to the higher concrete stress capacity of the cover and core concrete, hence reduced neutral axis 
depth, and is located within the what is called the primary hinge zone. This increase is over 
about half the depth of the column at each column end (where double-curvature occurs). 
Consequently, there is a greater moment demand just beyond this region, within the secondary 
hinge zone. Confinement enhancement, therefore, extends from the end of the column through 
the primary and secondary hinge zones, as shown in Figure 12.4.4-1A. Note that the 
categorization of primary and secondary hinge zones comes from Seible and Innamorato(1995). 

The confinement pressure also serves to clamp the lap splice connection of the column 
longitudinal bars. The thickness (effective clamping pressure) needed for lap splices is derived 
differently from the confinement requirements, however, as test results indicate that, at a dilation 
strain of about 0.001, lap splice slippage is initiated. These results, combined with the 
empirically derived radial pressure requirement to prevent slippage, determine the FRP overlay 
thickness. 

The FRP overlay thickness is determined for each of the three deficiencies; zones for these are 
shown in Figure 12.4.4-1A. As noted by Priestley, Seible, and Calvi (1996), the maximum of the 
three at any section should be provided; it should not be the sum, as reported in some other 
documents. This is because the lap-splice clamping action and compression concrete 
confinement occurs on opposite sides of the column, so the maximum requirement of the two 
will serve both well. Shear resistance of the FRP occurs along the column face parallel to shear 
load direction. The FRP anchors the concrete compression strut and is designed to maintain 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Figure 12.4.4-1A: Seismic Retrofit of Columns Using FRP Composites 

12-34 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Figure 12.4.4-1B: Seismic Retrofit of Columns Using FRP Composites 

aggregate interlock over the crack length. This contribution is a by-product of the hoop tension 
required for the confinement, so FRP composite thickness for shear need not be added to that 
required for confinement. 

With successful mitigation of the three deficiencies, column flexural hinges can be developed 
and deformation capacity will be increased. 

12-35 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Detailing Considerations 

The lap splices should be staggered, like that done with steel reinforcement, to mitigate a weak 
plane. As shown in Figure 12.4.4-1A, a nominal gap of about  between the FRP composite 
and the boundary elements (slab, beam or footing) is provided to prevent the overlay from 
bearing and, consequently, increasing the flexural capacity or stiffness of the column. 

Cost/Disruption 

To appropriately evaluate the cost of a retrofit scheme using and FRP overlay in comparison to 
traditional retrofit concepts (such as concrete or steel jacketing), one needs to consider the cost of 
the raw material, the level of specialization required by the contractor to install the system, the 
cost of labor and equipment, the cost of quality control and quality assurance, the temporary 
impact of disruption during construction, and the permanent impact to the building functions. 
Although FRP overlays are relatively expensive compared to steel and concrete, they can offer 
advantages when only limited access is available or minimal disruption of existing conditions is 
desired. 

Construction Considerations 

Access all around the column is often limited due to partition walls, ceilings and other 
architectural components, as well as structural elements of a building. Conditions at the base of 
the column in the lowest story must be carefully considered. The extent of FRP application 
normally will extend to the top of footing which may require local slab removal. The floor slab 
in the area may also interact with the column and affect strengthening requirements. This effect 
should be considered or a gap placed between the slab and column to prevent interaction. 

Proprietary Concerns 

See Section 13.4.1 for brief discussion of proprietary concerns. 

12.4.5 Enhance Concrete Column with Concrete or Steel Overlay 
Deficiencies Addressed by the Rehabilitation Technique 

Inadequate shear capacity 

Inadequate axial compression capacity 

Inadequate flexural plastic hinge confinement 

Inadequate lap splice 

Description of the Rehabilitation Technique 

Adding a fiber reinforced polymer (FRP) composite overlay to a concrete column is a recent 
approach to addressing seismic deficiencies, and is discussed in Section 12.4.4. Adding a 
concrete or steel jacket is a more traditional method of enhancing a deficient concrete column, 

Figure 12.4.5-1 provides examples of concrete and steel jacketing for a rectangular column. 
Because FRP overlays have become more common, this section will focus on FRP overlays. 

12-36 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Figure 12.4.5-1: Concrete and Steel Overlays for Concrete Columns 

12-37 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Design Considerations 

Research basis: Some research on concrete overlays is contained in FIB (2003); other related 
work is discussed in Section 12.4.5. Design of the overlay or jacket uses typical ACI 318 
concrete design principles. Sufficient drilled dowels between the overlay and existing concrete 
should be provided to achieve composite action. Research on steel overlays includes Engelhardt, 
et al., (1994). 

Detailing Considerations 

Concrete overlays: Concrete jacketing will take up a larger cross section than either FRP 
overlays or steel overlays. The surface of the existing concrete must be roughened appropriately. 
Reinforcing steel will need to be in at least two pieces to get it around the existing column. 
135 degree hooks are required for confining ties and may dictate the size of the overlay in order 
to provide enough room for the hook extension. 

Steel jackets: Steel jackets require at least two pieces to get around the existing column and 
involve field welding. Like FRP overlays, when the aspect ratio of a rectangular column gets too 
large, the jacket becomes less effective. The California Department of Transportation (Caltrans) 
uses elliptical jackets in these situations. The corners of the existing column will need to be 
trimmed so the steel can pass by. There is a gap between the steel and the concrete of at least  
that is filled with grout. A gap is also provided at the ends of the column to permit rotation 
without engagement in bearing of the steel jacket. 

Cost/Disruption 

To appropriately evaluate the cost of a retrofit scheme using and FRP overlay in comparison to 
traditional retrofit concepts (such as concrete or steel jacketing), one needs to consider the cost of 
the raw material, the level of specialization required by the contractor to install the system, the 
cost of labor and equipment, the cost of quality control and quality assurance, the temporary 
impact of disruption during construction, and the permanent impact to the building functions. 
Although FRP overlays are relatively expensive compared to steel and concrete, they can offer 
advantages when only limited access is available or minimal disruption of existing conditions is 
desired. 

Construction Considerations 

Concrete overlays: Because of the difficulty of placing an overlay on all sides of an existing 
column, concrete overlays are typically done with cast-in-place concrete, rather than shotcrete. 
The need for formwork is a significant disadvantage for concrete overlays, compared to FRP and 
steel overlays. Placing the concrete and vibrating are also challenging due to access limitations 
at the top of the column where it runs into beams or slabs. Pour ports or holes in the diaphragm 
are needed. 

Steel overlays: Steel overlays are typically 3/16 or  thick and become quite heavy. Access 
and lifting issues in existing buildings can force the overlay to be broken down into pieces, 
increasing the amount of field welding necessary to join the pieces back together. 

12-38 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Proprietary Concerns 

Unlike FRP overlays, no proprietary issues have been identified with using concrete of steel 
jackets. 

12.4.6 Enhance Concrete Moment Frame 
Deficiencies Addressed by the Rehabilitation Technique 

Inadequate global shear capacity 

Inadequate lateral displacement (global stiffness) capacity 

Inadequate ductile detailing for shear strength, confinement, or strong column/weak beam 

Description of the Rehabilitation Technique 

An alternative method to either add strength and stiffness to an existing concrete moment frame 
or correct non-ductile detailing deficiencies of the frame members is by direct enhancement: 
increasing the size of the columns and beams of the frame with new reinforced concrete. This 
method entails adding a jacket of reinforced concrete around the existing columns and beams, an 
approach similar to jacketing by steel or fiber wrap. The new concrete may be either cast-inplace 
or shotcrete. This approach is relatively rarely employed in the U.S. because it is labor 
intensive, but when existing openings must be maintained and walls or braced frames cannot be 
installed, it may be rehabilitation technique worth considering. 

Design Considerations 

Research basis: Alcocer and Jirsa (1991) tested reinforced concrete beam-column connections 
subjected to unidirectional and bidirectional cyclic loading up to displacements equivalent to 4% 
interstory drift. Jacketing of columns resulted in strong column  weak beam behavior, with 
increases in peak strength of over four times the existing assembly. Jacketing of beams as well 
as columns improved joint confinement, decreased stresses on existing beam reinforcement, and 
provided some additional strength increases. The use of jacketing to repair damaged frames was 
also tested. When comparing the performance of a specimen in which a damaged column and 
joint were repaired by jacketing to that of a similar specimen in which the original column and 
joint were undamaged, the damaged and repaired assembly exhibited 65% of the strength at 2% 
drift and 50% of the stiffness at 0.5% drift relative to the undamaged retrofitted specimen. 

Reinhorn, Bracci, and Mander (1993) performed shake table tests on a one-third scale model of a 
three-story, one-bay by three-bays, concrete moment frame. The tests compared the 
performance of an unretrofitted building, subjected to peak ground acceleration 0.30 g, to that of 
a similar structure retrofitted by jacketing interior columns, post-tensioning added longitudinal 
column reinforcement, and providing a reinforced concrete fillet infill around beam-column 
joints. These retrofit measures were intended to ensure strong column - weak beam behavior, 
enhance joint shear capacity, and improve anchorage for discontinuous beam reinforcement. 
Tests demonstrated that the retrofitted structure exhibited significantly reduced column damage, 
especially in the first story, and improved ductility in yielding beams; however, lateral 
displacements remained quite large, reaching a maximum inter-story drift of 2.1% at the 
first floor. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Impact on architectural and M/E/P systems and components: Enlarging columns, beams, or 
joints may impact existing ceilings, partitions, ductwork, plumbing, lighting, or usable floor 
space. The designer should keep in mind that the space required for access, formwork, and 
finishes will be greater than the final dimensions of the jacketed member. 

System design: New transverse and longitudinal reinforcement should be designed such that 
members are flexurally-governed and beams yield before columns do. The jacketing of the 
frame will result in increases in system stiffness that may result in increased attracted load. Both 
new and existing concrete should be considered in developing composite properties for modeling 
and design. Joint shear can be the weak link in existing moment frames. A significant focus of 
the Alcocer and Jirsa (1991) research was to provide special angles at the joint region to improve 
joint confinement. Stiffness at 0.5% drift was well predicted when cracked properties of 0.5EI 
were used for the beams and 0.5EI were used for the columns. Stiffness decreased about 40% to 
50% as drifts increased from 0.5% to 2.0%. 

Detailing Considerations 

Surface preparation: In order for a new reinforced concrete jacket to act compositely with an 
existing member, sufficient bond must exist between the new and existing concrete. The Alcocer 
and Jirsa (1991) tests made use of findings from Bass, Carrasquillo, and Jirsa (1985). In the 
Alcocer and Jirsa (1991) tests, they used a handheld electric chipping hammer to reveal some 
aggregate. The amplitude is not documented in the report, but the implication is that it was less 
than the traditional  amplitude value. Dust was cleaned with a thick brush and vacuum 
cleaner. A bonding agent was not used, and the existing concrete surfaces were not saturated in 
all cases. Bonding agents in current practice are not common, but some engineers do 
recommend prewetting the existing concrete. 

Column jacketing: Beams are typically the same plan dimension or narrower that the supporting 
column. New column bars can then pass by the existing beam bars through holes cut in the floor 
slab. Alcocer and Jirsa (1991) explored distributed longitudinal bars vs. bundled bars in the 
corners of the new jacket, but did not find significant differences. It is not possible to put a one-
piece closed hoop around an existing column. One approach is to place a U-bar on three sides 
with a single leg closure piece, such as shown in Figure 12.4.4-2A. Alcocer and Jirsa (1991) 
used two overlapping L-shaped bars, each with 135-degree hooks at the ends. A minimum 
column jacket thickness of 4 was recommended. 

Beam jacketing: When beams are jacketed, the bottom and sides can be increased in dimension, 
but typically the top of beam must remain at the existing top of slab level. In the Alcocer and 
Jirsa (1991) tests, slots were cut in the top of the slab just above the beam with holes at the ends 
of the slot through slab. Inverted U-bars were placed in the slot and through the holes. The 
bottom of the inverted U-bars overlapped around beam longitudinal steel with U bars 
surrounding the beam. A minimum beam jacket thickness of 3 is recommended. Care must be 
taken to investigate existing beam reinforcing, so that appropriate locations for new beam 
longitudinal bars can be located to miss existing reinforcing in orthogonal beams and in columns. 

Joint enhancement: In the Alcocer and Jirsa tests (1991), vertical steel angles were placed in 
holes cut through the slab at the four corners of the column and then welded to horizontal steel 

12-40 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

bars above the slab and below the beam, to form a cubical cage. In the Reinhorn, Bracci, and 
Mander (1993) tests, joint strength was enhanced by infilling between orthogonal beams with 
concrete to create a diamond-shaped fillet in plan. Reinforcing at the outer ends of the fillet 
was placed through holes cut into the beam just under the slab and at the base of the beam. 

Cost/Disruption 

The cost and disruption associated with jacketing a concrete frame is significant because it 
potentially involves complicated formwork, reinforcement, and concrete pouring over much of 
the building, as opposed to more common shear wall approaches where the walls are only placed 
in localized areas. Beam jacketing is much more invasive and time consuming and is typically 
less important than the column jacketing. Alcocer and Jirsa (1991) reported that the construction 
time required for jacketing beams and columns was nine times that required for jacketing 
columns alone. 

Construction Considerations 

Noise and disruption: Removing existing column, beam, ceiling and floor finishes is disruptive. 
Chipping the cover concrete off of existing frame members is noisy and disruptive to occupants, 
as is cutting holes in the floor slab. As such this particular technique can be less desirable than 
many others. 

Mix design: Due to the narrow thickness of the jacket and difficult working conditions, the mix 
design should emphasize ease of placement, by using small size aggregate and water-reducing 
admixtures. In the Alcocer and Jirsa (1991) tests, they used 3/8 maximum size aggregate and 
superplasticizer. 

Concrete placement: Concrete should be placed from above through holes in the slab for both 
columns and slabs. Holes have to be big enough for both the concrete hose and the vibrator. 
Consideration can be given to leaving gaps at the top of the forms on the sides of the beams 
where they meet the slabs for air relief vents. 

Proprietary Concerns 

The basic materials are generic. 

12.5 References 
ACI 440.2R-02, 2002, Guide for the Design and Construction of Externally Bonded FRP 
Systems for Strengthening Concrete Structures, American Concrete Institute, Chicago, IL. 

Alcocer, S.M. and J.O. Jirsa, 1991, Reinforced Concrete Frame Connections Rehabilitated by 
Jacketing, Phil M. Ferguson Structural Engineering Laboratory Report 91-1, University of Texas 
at Austin, Austin, TX, July. 

Bass, R.A., R.L. Carrasquillo, and J.O. Jirsa, 1985, Interface Shear Capacity of Concrete 
Surfaces Used in Strengthening Structures, Phil M. Ferguson Structural Engineering Laboratory 
Report 85-4, University of Texas at Austin, Austin, TX, December. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 12  Type C1: Concrete Moment Frames 

Engelhardt, M.D., et al., 1994, "Strengthening and Repair of Nonductile Reinforced Concrete 
Frames Using External Steel Jackets and Plates," Repair and Rehabilitation Research for Seismic 
Resistance of Structures, ed. James Jirsa, Report R/R 1994-1, June. 

FIB (Federation Internationale de Beton), 2003, Seismic Assessment and Retrofit of Reinforced 
Concrete Buildings, State-of-Art Report, Bulletin 24. 

Jimnez-Pacheco, J. and M.E. Kreger, 1993, Behavior of Steel-to-Concrete Connections for Use 
in Repair and Rehabilitation of Reinforced Concrete Structures, Phil M. Ferguson Structural 
Engineering Laboratory Report 93-2, University of Texas at Austin, Austin, TX, March. 

Jones, E.A. and J.O. Jirsa, 1986, Seismic Strengthening of a Reinforced Concrete Frame Using 
Structural Steel Bracing, PMFSEL Report No. 86-5, Department of Civil Engineering/Bureau of 
Engineering Research, University of Texas, Austin, TX. 

Priestley, N., Seible, F. and G.M. Calvi, 1996, Seismic Design and Retrofit of Bridges, John 
Wiley & Sons, New York, NY. 

Reinhorn, A.M., J.M. Bracci, and J.B. Mander, 1993, Seismic Retrofit of Gravity Load 
Designed Reinforced Concrete Buildings, Proceedings of the 1993 National Earthquake 
Conference: Earthquake Hazard Reduction in the Central and Eastern United States: A Time for 
Examination and Action, Memphis, TN, May, Vol. 2, pp. 245-254. 

Seible, F. and D. Innamorato, 1995, Earthquake Retrofit of Bridge Columns with Continuous 
Carbon Fiber Jackets, Volume II, Design Guidelines, Report to Caltrans, Division of 
Structures, University of California, San Diego, La Jolla, CA. 

12-42 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 13  Type C2b: Concrete Shear Walls 

Chapter 13 - Building Type C2b: Concrete Shear Walls 
(Bearing Wall Systems) 

13.1 Description of the Model Building Type 
Reinforced concrete walls in a building will act as shear walls whether designed for that purpose 
or not. Therefore, cast-in-place concrete buildings that contain any significant amount of 
concrete wall will fall into this category. However, there are two distinctly different types of 
concrete wall buildings, those that contain an essentially complete beam/slab and column gravity 
system, and those that use bearing walls to support gravity load and have only incidental beam 
and column framing. In this document, these building types have been separated and are 
designated C2f for the gravity frame system and C2b for the bearing wall. This section covers 
the bearing wall type. In this type of building, all walls usually act as both bearing and shear 
walls. The building type is similar and often used in the same occupancies as Building Type 
RM2, namely in mid- and low-rise hotels and motels. This system is also used in residential 
apartment/condominium type buildings. 

Figure 13.1-1: Building Type C2b: Concrete Shear Walls (Bearing Wall Systems) 

13-1 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 13  Type C2b: Concrete Shear Walls 

Variations Within the Building Type 

In order for this framing system to be efficient, a regular and repeating pattern of concrete walls 
are required to provide support points for the floor framing. In addition, since it is difficult and 
expensive to make significant changes in the plan during the life of the building, planning 
flexibility is not normally an important characteristic when this structural system is employed. 
The occupancy type that most often fit these characteristics are residential buildings, including 
dormitories, apartments, motels, and hotels. These buildings will often be configured with 
reinforced concrete bearing walls between roomsalso acting as shear walls in the transverse 
direction, and reinforced concrete walls on the interior corridor acting primarily as shear walls in 
the longitudinal direction. Sometimes the longitudinal lateral system includes the exterior wall 
system, although this wall is normally made as open as possible. In any case, the wide variation 
in structural layouts and occupancies that is included in Building Type C2f is not seen in 
Type C2b. 

It is seldom possible to plan a building layout that provides complete gravity support with walls. 
Often, local areas are supported with isolated columns, and sometimes beams and girders are 
also necessary, but story heights in these buildings are usually small and added depth in the floor 
framing system is difficult to obtain. The extent of such beam and column framing often causes 
confusion between Building Types C2b and C2f, but buildings should have an essentially 
complete gravity frame system to be placed in C2f. If significant plan area is supported solely 
by walls, the structures are normally classified as C2b. 

There are important variations in floor framing systems employed in this building type, and their 
adequacy to act as a diaphragm is an important characteristic of this building type as discussed 
below. 

Floor and Roof Diaphragms 

The parallel layouts of supporting walls and the need to minimize story heights normally leads to 
the use of one-way uniform-depth concrete floor systems. Cast-in-place and precast systems, 
both conventionally reinforced and prestressed, have been employed. The precast systems are 
often built up of narrow planks, which may not provide an adequate diaphragm unless a cast-inplace 
topping is provided. In addition, the precast systems may be placed with only a very 
narrow bearing area on the supporting walls, which may be inadequate to provide vertical 
support during seismic movements. The adequacy of the shear connection between slab and 
walls is also often an issue for both cast-in-place and precast systems. 

Foundations 

The bearing walls obviously require some kind of starter beam at grade for construction purposes 
and this often leads to a simple continuous grade beam system. In poor soils, piles or drilled 
piers may be added below the grade beam. A continuous mat foundation may also be employed 
due to the short spans and total length of bearing points in this building type. 

13-2 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 13  Type C2b: Concrete Shear Walls 

13.2 Seismic Response Characteristics 
Due to the extent of wall, bearing wall buildings will be quite stiff. Elastic and early post-elastic 
response will therefore be characterized with lower-than average drifts and higher-than-average 
floor accelerations. Damage in this range of response should be minimal. 

Overall post-elastic response may often include rocking at the foundation level. If rocking does 
not occur, the height-to-length ratio of shear walls in these buildings may force shear yielding 
near the base, which may lead to strength and stiffness degradation. 

Global stability may also be compromised by poor connections between floor slab construction 
and bearing walls. 

Shear Wall Behavior 

When subjected to ever increasing lateral load, individual shear walls or piers will first often 
force yielding in spandrels, slabs, or other horizontal components restricting their drift, and 
eventually either rock on their foundations, suffer shear cracking and yielding, or form a flexural 
hinge near the base. Shear and flexural behavior is quite different, and estimates of the 
controlling action are affected by the distribution of lateral loads over the height of the structure. 

Yielding of spandrels, slabs, or other coupling beams can cause a significant loss of stiffness in 
the structure. Flexural yielding will tend to maintain the strength of the system, but shear 
yielding, unless well detailed, will degrade the strength of the coupling component and the 
individual shear wall or pier will begin to act as a cantilever from its base. In this building type, 
the coupling elements are often slabs, and their lack of bending stiffness may reduce or eliminate 
significant coupling action. 

Rocking is often beneficial, limiting the response of the superstructure. However, the amplified 
drift in the superstructure from rocking must be considered. In addition, if varying wall lengths 
or different foundation conditions lead to isolated or sequencing rocking, the transfer of load 
from rocking walls must be investigated. In buildings with basements, the couple created from 
horizontal restraint at the ground floor diaphragm and the basement floor/foundation (sometimes 
called the backstay effect) may be stiffer and stronger than the rocking restraint at the 
foundation and should be considered in those configurations. 

Shear cracking and yielding of the wall itself is generally considered undesirable, because the 
strength and stiffness will quickly degrade, increasing drifts in general, as well as potentially 
creating a soft story or torsional response. However, in accordance with FEMA 356, shear 
yielding walls or systems can be shown to be adequate for small target displacements. Type C2b 
buildings will often fall into this category. 

Flexural hinging is considered ductile in FEMA 356 and will degrade the strength of the wall 
only for larger drifts. Similar to rocking, the global effect of the loss of stiffness of a hinging 
wall must be investigated. 

13-3 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 13  Type C2b: Concrete Shear Walls 

13.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
See Table 13.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 
Selected deficiencies are further discussed below by category. 

Global Strength 

Due to the extensive use of walls, buildings of this type seldom have deficiencies in this 
category, unless significant degradation of strength occurs due to shear failures. 

Global Stiffness 

Similar to strength, global stiffness is seldom a problem in this building type. However, the 
effect of coupling slabs on initial stiffness and the potential change in stiffness due to yielding of 
these coupling slabs or wall-beams over doors should be investigated. 

Configuration 

The most common configuration deficiencies in this building type are weak or soft stories 
created by walls that change configuration or are eliminated at the lower floors. It is difficult to 
provide the needed ductility at the weak story, and often strength must be added. Completely 
discontinuous walls also create a load transfer deficiency for both overturning and shear. In such 
cases, collectors are often needed in the floor diaphragm, and supporting columns need axial 
strengthening. 

Load Path 

A common deficiency in this building is weakness in the load path from floor to walls, either 
collector weaknesses or shear transfer weakness immediately at the floor wall interface. Local 
transfer can be strengthened by adding concrete or steel corbel elements, dowels, or 
combinations of these components. As indicated above discontinuous walls also often create 
load path deficiencies. 

Component Detailing 

The most common detailing problem in this building type is an imbalance of shear and flexural 
strength in the walls, leading to pre-emptive shear failure. This deficiency may be shown to be 
not critical with small displacement demands, walls can be strengthened in shear with overlays 
of concrete, steel, or FRP. 

The layout of walls often forces coupling between walls through the slab system or across 
headers of vertically aligned doors. These coupling components are seldom designed for the 
coupling distortions that they will undergo, particularly in older buildings. Short lengths of slabs 
between adjacent walls receive damage by coupling action that could compromise the gravity 
capacity. It is difficult to add strength or ductility to these slab areas, but vertical support at 
support points can be supplemented by corbels of steel or concrete. Damage to headers over 
doors often does not contribute to deterioration of overall response and can sometimes be 
acceptable. Local areas of wall can also be strengthened by overlays of concrete, steel, or FRP. 

13-4 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 13  Type C2b: Concrete Shear Walls 

Table 13.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2b Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Global 
Strength 
Insufficient in-
plane wall shear 
strength 
Concrete/masonry 
shear wall [12.4.2] 
Concrete wall overlay 
[21.4.5] 
Fiber composite wall 
overlay [13.4.1] 
Steel overlay 
Seismic isolation 
[24.3] 
Reduce flexural 
capacity [13.4.4] 
Insufficient 
flexural capacity 
Concrete/masonry 
shear wall [12.4.2] 
Add chords [12.4.3] 
Inadequate 
capacity of 
coupling beams 
Concrete/masonry 
shear wall [12.4.2] 
Strengthen beams 
[12.4.2] 
Improve ductility of 
beams [12.4.2] 
Remove beams 
Global 
Stiffness 
Excess drift 
(normally near the 
top of the 
building) 
Concrete/masonry 
shear wall [12.4.2] 
Fiber composite wrap 
of columns to improve 
lateral displacement 
capability [12.4.4] 
Provide detailing of all 
other elements to 
accept drifts 
Concrete wall overlay 
[21.4.5] 
Supplemental 
damping [24.4] 
Configuration Discontinuous 
walls 
Add wall or adequate 
columns beneath 
[12.4.2] 
Fiber composite wrap 
of supporting columns 
[12.4.4] 
Concrete/steel jacket of 
supporting columns 
[12.4.5] 
Improve connection to 
diaphragm [13.4.3] 
Remove wall 
Soft story or weak 
story 
Add strength or 
stiffness in story to 
match balance of 
floors 

13-5 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 13  Type C2b: Concrete Shear Walls 

Table 13.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2b Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Configuration 
(continued) 
Re-entrant corner Add floor area to 
minimize effect of 
corner 
Provide chords in 
diaphragm [12.4.3] 
Torsional layout Add balancing walls 
[12.4.2] 
Load Path Inadequate 
collector 
Add steel or concrete 
collector [12.4.3] 
Inadequate slab 
bearing on walls 
Add diagonal dowels 
[13.4.3] 
Add steel ledger 
[13.4.3] 
Component 
Detailing 
Wall inadequate 
for out-of-plane 
bending 
Add strongbacks 
[21.4.3] 
Concrete wall overlay 
[21.4.5] 
Wall shear critical Concrete wall overlay 
[21.4.5] 
Fiber composite wall 
overlay [13.4.1] 
Reduce flexural 
capacity [13.4.4] 
Diaphragms Precast 
components 
without topping 
Improve 
interconnection 
[22.2.11] 
Add topping 
Inadequate in-
plane shear 
capacity 
R/C topping slab 
overlay 
Fiber composite 
overlay [22.2.5] 
Inadequate shear 
transfer to walls 
Add diagonal drilled 
dowels [13.4.3] 
Add steel angle ledger 
[13.4.3] 

13-6 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 13  Type C2b: Concrete Shear Walls 

Table 13.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2b Buildings 
Deficiency Rehabilitation Technique 
Category 
Diaphragms 
(continued) 
Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Inadequate chord 
capacity 
New concrete or 
steel chord member 
[12.4.3] 
Excessive stresses 
at openings and 
irregularities 
Add chords [12.4.3] Infill openings 
[22.2.4] 
Foundations See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 

13-7 


Techniques for the Seismic Rehabilitation 
of Existing Buildings: FEMA 547 Chapter 13  Type C2b: Concrete Shear Walls 

Diaphragm Deficiencies 

Precast floor systems used in this building type often provide inadequate diaphragm behavior 
that could lead to bearing failures at the floor wall interface, particularly when no topping slab is 
present. Some topping slabs used primarily for leveling and smoothing the floor are 
inadequately tied to the precast elements or the walls, and are too thin or poorly reinforced to act 
as diaphragms on their own. See Chapter 22. 

Foundation Deficiencies 

This building type often places large demands on the foundation system. If rocking is shown to 
be a controlling displacement fuse for the building, the foundations must be investigated to 
assure that these displacements can safely occur. See Chapter 23. 

13.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
13.4.1 Enhance Shear Wall with Fiber-Reinforced Polymer Composite Overlay 
Fiber-Reinforced Polymer Composite Overview 

In addition to this section, seismic rehabilitation techniques using fiber-reinforced polymers are 
described in several other sections in this document, including Section 12.4.4, Enhance Column 
with Fiber-reinforced Polymer Composite Overlay, and Section 22.2.5, Enhance Slab with 
Fiber Reinforced Polymer Composite Overlay. This section provides a general overview of FRP 
characteristics. 

Composite makeup and application: The construction industrys term Fiber-Reinforced Polymer 
(FRP) refers to a composite material made up of carbon, fiberglass or aramid (Kevlar) fibers that 
are bound together by either a resin or ester polymer. These are commonly referred to as glass 
fiber-reinforced polymer (GFRP) composite, carbon fiber-reinforced polymer (CFRP) 
composite, or aramid fiber-reinforced polymer (AFRP) composite, respectively. The raw fibers 
(synonymous to filaments) can be woven to form mesh (uni- or bi-directional with the 
orientation of the warp to the weft at 45 or 90 degrees), collected together to form a carbon tow 
winding (an untwisted bundle of continuous filaments) or sheet, or pultruded to form a 
prefabricated plate or other shape. 

Usually, at least two layers of FRP are applied to the exterior concrete (substrate) surface. For 
beam and column applications, one layer should be considered sacrificial, due to the possibility 
of abrasion and the fact that lap splices are used and potentially compromise a layers 
effectiveness. 

Mechanical properties: The mechanical property in the direction of the fibers of GFRP, CFRP 
and AFRP is an essentially linear-elastic response followed by sudden rupture. The three 
material systems have different rupture strains and moduli. This variation becomes an important 
consideration when selecting the fiber type. While the rupture strain of each material is 
different, they are all significantly less (by approximately an order of magnitude) than that of 
conventional concrete reinforcement, which leads to compatibility issues. 

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Creep (increase in strain), stress rupture (reduced tensile capacity) and stress corrosion 
(corrosion that is dependent on presence of stress to occur) are phenomena that occur when FRP 
is subjected to sustained loads, such as flexural strengthening of slabs for long-term gravity 
loads. Each type of composite responds differently to loading regime, environmental conditions, 
and matrix and fiber make-up: AFRP is prone to creep under sustained load and moderately so to 
stress corrosion, GFRP is prone to stress rupture (as low as 20% of the ultimate), but CFRP is 
robust to sustained loads. Page 115 of FIB (2001) reference provides more information. This 
document addresses seismic loading, which is of short duration; hence, FRP is not affected by 
these characteristics. Accidental sustained stress loading for seismic applications may occur and 
should be given special study. An example is the FRP diaphragm overlays where the fiber 
composite transfer on top of a floor to a wall can act as negative slab reinforcement and resist 
subsequent live loads. Gravity load enhancement is not addressed in this document, but it should 
be noted that it is affected by this characteristic. 

For the same FRP composite, mechanical properties can vary between manufacturers and 
sometimes within the same manufacturer over the course of material production for a project. It 
is recommended that the contract documents clearly specify the performance requirements (such 
as force per unit width for each application) and the minimum ultimate rupture strain. 

Requirements at the FRP-to-substrate interface: There are contact-critical and bond-critical 
applications between the composite material and the substrate surface. 

Contact-critical applications are mostly limited to beam and column shear and confinement 
enhancement techniques, and it is preferred that the composite be wrapped around all sides of the 
element (i.e., made continuous). These applications do not require shear flow capacity between 
the composite to the substrate, so paint and other smooth finish materials may be left in place. 

Bond-critical applications, such as discontinuous applications and wall applications, require 
shear flow capacity between the FRP overlay and the substrate. (Note that flexural strengthening 
for gravity load enhancement of slabs and beams is a bond-critical application, but it is not 
included in this document.) The surface preparation is important and a concrete surface profile of 
3 (CSP 3) is required (as described in ICRI, 2003), which calls for all loose laitance, the weaker 
outer cement paste layer, and any unsound concrete to be removed. Light sand or water blasting 
readily achieves the desired result. The surface should be dust free at the time of applying the 
first resin layer. Resin putty is used to fill the voids, provide a smooth surface, and create a 
chamfered corner. Paint and other surface finishes must be checked for hazardous materials 
before removal. 

All of the fiber composites have the inherent tendency to rupture prematurely at stress 
concentrations. Such concentrations are formed by burrs, sharp edges, protrusions, etc., in the 
substrate. Any sharp edge, or protrusion of any kind, must be removed during the preparation of 
the substrate to ensure a smooth surface free of dirt, grease, oil and finishes. Existing elements 
damaged by cracks or corrosion should be repaired prior to applying this rehabilitation 
technique. 

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The substrate surface should be essentially flat, so that the fibers are straight when positioned. 
An uneven surface prevents the fibers from lying straight, and upon loading, will tend to 
straighten, which will compromise the bond capacity. An out-of-plane angle of 1-2% will 
compromise the bond strength by initiating peeling of the FRP overlay from the substrate. 

Typically, the weak link in bond-critical applications is the tensile rupture of the substrate 
(although the resin should also be checked). This weak link obtains peak strength at about 1/32inch 
displacement, followed by complete loss of bond strength at about 1/3-inch. Further, test 
results and observed behavior show a finite development length of about a few inches and 
beyond that no additional load is developed in the composite material, meaning that the fiber 
material does not necessarily develop its full strength (see Teng et al. (2002) Figures 2.3 and 
2.7). This behavior is unlike rebar reinforcement, where the bar development length is sized, and 
the bond area is provided to attain full rebar strength. The design, therefore, for FRP composite 
requires focus on bond strength and an awareness of crack patterns and the locations of inelastic 
deformations. 

Fiber and mechanical anchors: Significant research on new means of anchoring FRP composite 
to a substrate is occurring at this time. The most common anchor consists of inserting carbon 
fiber or a mixture of carbon and glass fibers into a drilled hole in the substrate. The hole is then 
filled with epoxy, and the protruding strands are splayed to a cone or fan shape that is used to lap 
with the composite overlay. The splay is located between the overlay layers to enhance the 
splice connection. There are many variations of this approach being studied. Anchor spacing, 
edge distance and other issues that influence the performance are also being investigated. The 
design of the anchor and FRP composite overlay system, therefore, should be case specific and 
based on the most recent research. 

Durability: Composite materials, if manufactured correctly and with the appropriate finish 
applied, can provide corrosion resistance, ultraviolet light resistance, fire resistance, and 
tolerance to variations in temperature  making them generally suitable for most environments. 
Specific consideration should be given to galvanic corrosion of CFRP, particularly where there is 
an electrolyte present and potential for mixing of metals. Although CFRP is otherwise corrosion 
resistant, it is a conductor and will change properties when heated. Locations where lightning 
strikes may occur, such as garage rooftops or exterior of buildings, will warrant the use of other 
FRP materials or grounding of metal grid to protect the CFRP. See FIB (2001), Section 9.11. 
For each environment application, the experience of manufacturers and researchers should be 
considered, and the manufacturers warranties should be carefully considered. The designer 
must consider the myriad of environmental factors and develop an FRP solution that is 
appropriate for those conditions, and, more importantly ensure that the appropriate fiber and 
polymer material, surface protection, and finish is specified. As an example, where high 
humidity and/or high alkalinity are present, carbon fiber is the preferred choice. Where 
ultraviolet light is present, surface protection of the composite can be achieved with an acrylic or 
polyurethane based paint applied when the polymer is still tacky. For more durability 
information, refer to FIB (2001), Chapter 9. 

Constructibility: In addition to the costs of material and installation of FRP composite, concrete 
surface preparation and final appearance requirements must also be considered. The surface 

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finish requirements, as described above in the Interface Requirements section, need to be 
included and made clear in the contract documents. The woven texture of the mat applications 
does read through to the finish surface. If additional surface finishing is required, such as paint 
or a cementitous appearance, this can add significant cost, and should be coordinated with the 
architect and owner. For seismic rehabilitation applications, the FRP overlay is similar to 
structural steel braces, in that they are not typically required to support gravity loads and thus do 
not require fireproofing. However, some local jurisdictions do require that a fire-protecting 
surface be provided, such as intumescent paint. 

The resins shelf-life, pot-life, ambient temperature, ventilation, substrate moisture state and 
other environmental factors need to be carefully considered by the design team. For example, 
due to issues with offgassing during and following installation for some time, not only will 
ventilation be needed, but occupants may need to be temporarily displaced. Quality control 
procedures, which should be required by the contract documents, need to address these issues 
and be verified as acceptable by the engineer of the project and the inspector of record. 

Research basis: Teng, et al. (2002) provides a summary of recent research projects and design 
equations suggested by researchers. Design equations and construction quality assurance 
requirements are also present in model codes, such as ACI 440.2R-02 (2002) and FIB 14 (2001). 
For a summary of the mechanical properties and design procedures, see Priestley, et al. (1996). 
For detailed material mechanical properties and chemical background information, see Kaw 
(1997). 

Deficiency Addressed by the Rehabilitation Technique 

Inadequate shear capacity in a concrete shear wall 

Description of the Rehabilitation Technique 

An FRP overlay is a technique that is used to enhance the in-plane shear capacity of a reinforced 
concrete shear wall. The overlay can be applied to one or both sides of a wall, and where 
possible, should wrap around the ends of the wall to aid in anchoring the overlay. The 
rehabilitation technique is bond-critical, regardless of whether or not the material wraps around 
the end of the walls. Uni-directional (horizontal-oriented) fibers are used to enhance the shear 
capacity, creating a predominately flexural post-yield response. Vertically-oriented fibers in bidirectional 
layouts will limit the vertical strains to that of the FRP composite, inhibiting the 
ductile behavior. Therefore, the rehabilitation technique for walls is limited to horizontally 
oriented fibers, unless there are extenuating circumstances. The shear enhancement may change 
the walls response from a shear-dominated behavior to a flexural, sliding shear, or rocking 
behavior. At coupling beams, vertically-oriented fibers are typically used. See Figure 13.4.1.1A 
for examples of wall and coupling beam layouts. 

Design Considerations 

Research basis: For wall-specific criteria, there has been an increasing amount of research on 
strengthening of unreinforced masonry walls, but less information on reinforced concrete walls. 
Ghobarah (2004) is one source; and Laursen, Seible, and Hegemier (1995), though based on 
CMU specimens, can be extended to some degree to reinforced concrete walls. 

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Figure 13.4.1-1A: Shear Strengthening of Concrete Shear Walls Using FRP Composite 

The shear resistance contribution from the FRP is obtained in a similar manner to that used for 
wall reinforcement; the horizontal bars and FRP resist the horizontal shear (see Section 5.1 in 
Laursen, Seible, and Hegemier, 1995). The effective fiber area per unit width, and its 
contribution to shear resistance, is limited to the bond and anchorage strength capacity; providing 
additional fiber area will not provide additional shear capacity. Testing has been limited to 
mostly single and double layers of FRP per side of wall. Wrapping the material around the ends 
of the wall and/or providing fiber anchors will enhance the effectiveness of the overlay, 
particularly where cracks form, by increasing the anchorage and bond of the FRP to the substrate 

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Figure 13.4.1-1B: Fiber Anchor Details 

In terms of wall shear strength, the FRP overlay can be additive to the walls concrete and 
reinforcement contribution. The ductility, however, is dependent on the type of governing 
mechanism. For shear-dominated walls (where the post-yield deformations require slippage at 
crack locations), the FRP has limited ability to accommodate such deformation; hence, there is 
likely very limited ductility. Conversely, a wall dominated by flexural yielding is able to 
accommodate the plastic deformations by the yielding of the wall vertical reinforcement, 
provided that the wall vertical reinforcement is developed. As a result, the typical goal in adding 
an FRP overlay is to make a wall flexurally-critical. For the case where flexural yielding occurs 
first, and is then followed by shear yielding as a result of the reduced concrete shear contribution 
(due to the reduced aggregate interlock effectiveness), ductility is significantly less than that with 
a flexural yielding response. Moreover, there is essentially no confining pressure afforded by the 
use of FRP overlay, except locally, where a fiber anchor is used. Therefore, lap splice 

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performance is not enhanced with this technique and, depending on the size of bar, cover, and 
lap length, limited ductility may result. 

Although testing with bi-directional (at plus and minus 45 degrees) fibers in wall overlays shows 
enhanced shear and flexural strengths and moderately enhanced displacement ductility capacity, 
the vertical force component of the fibers contribution to the flexural strength may have adverse 
affects. This may reduce the distribution of inelastic straining and/or change the strength 
hierarchy to that of shear controlled. Until further research is performed, horizontally-oriented 
FRP strengthening is recommended. 

Detailing Considerations 

Given the high dependence on the bond strength of the FRP overlay to the substrate, in-situ bond 
testing should be included as a requirement in the contract documents. A testing program will 
verify the design assumptions and assist in providing quality assurance. If pilasters are present, 
either within the wall length or at wall ends, installing fiber anchors or removing portions of the 
pilaster should be considered to enhance the anchoring of the FRP overlay. 

Construction Considerations 

As discussed in the FRP Composite Overview section earlier, the engineer should inspect the 
surface of the wall elements to be rehabilitated and note in the contract documents the surface 
condition and wall configuration (e.g., wall corner profile and wall-to-slab configuration). To 
aid in developing a sound bid price, the contract documents need to record the as-built condition, 
including surface anomalies and configuration, and the surface preparation requirements. If the 
surface has been board formed with wood planks, for example, calling this out in the 
construction documents will enable a more accurate bid. Bid documents should also require the 
contractor to view existing conditions before bidding. 

Proprietary Concerns 

Although the basic materials are generic, the fabric that will be supplied is proprietary. Strand 
orientation and density, epoxy overlays, preparation requirements, and installation procedures 
may be different between suppliers. Some suppliers may not have experienced applicators in the 
area of the project. These variations must be considered to achieve an adequate specification, 
particularly if competition between suppliers is desirable. 

13.4.2 Enhance Deficient Coupling Beam or Slab 
Deficiencies Addressed by the Rehabilitation Technique 

Inadequate shear and bending capacity of coupling beams or slabs 

Inadequate ductile detailing for shear strength 

Description of the Rehabilitation Technique 

Coupling beam deficiencies can be encountered in any type of building structure that has shear 
walls for their primary lateral force-resisting system. In a bearing wall structure, however, the 
coupling beam is often not a beam at all but only the relatively thin concrete slab linking the 
adjacent walls across a corridor. In some cases, particularly near the top of taller buildings, the 

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restraint of a slab or beam in-line with the flexural deformed shape of a narrow shear wall can 
create coupling beam type issues at the tip of a single wall. In many cases, the slab consists of 
precast (and prestressed) hollow core planks, with or without a topping, with limited reinforcing 
steel at the link beam-wall joint instead of cast-in-place concrete. Furthermore, because bearing 
wall buildings are often residential buildings with short story heights, there is often no ability to 
create a dropped beam, and it is very difficult or impossible to increase the strength or ductility 
of the linking slab-beam itself. In these cases, the mitigation approach is to install a steel or 
concrete corbel to provide supplemental vertical support. 

For those cases where there actually is a beam or deeper header linking the walls, the most 
common mitigation method to strengthen it or to correct its non-ductile detailing deficiencies is 
by direct enhancement with new added reinforced concrete. The new concrete is generally added 
to one face of the coupling beam, in concert with concrete strengthening of the adjacent shear 
walls as well. In some cases, new reinforced concrete can be added to both faces and extended 
along the walls for development. In addition, or as an alternative, it may be possible to install a 
limited number of diagonal dowels to enhance the shear capacity of the coupling beam to wall 
joint. In a case where there is enough beam or header depth, two groups of tied bars could be 
placed in an X-configuration, embedded within the new beam reinforcement cage. The new 
concrete may be either cast-in-place or shotcrete. 

Design Considerations 

If the restraint of the slabs and beams described above is not needed for overall structural 
stiffness, the element should be evaluated for gravity support in a damaged state. No mitigation 
may be required, particularly in cases where cast-in-place concrete slabs exist. For cases where 
gravity support is inadequate, the approach may be the installation of a supplemental support in 
the form of new corbels placed under locations of damage. If there is no ceiling, care should be 
taken to minimize the visual impact of the corbel and connectors. If existing M/E/P system 
components conflict with installation of these supplemental supports, they should be relocated as 
required to make this minimum mitigation effort. 

For cases where an actual coupling beam or header is being augmented, the contribution of both 
the existing and new portions of the now composite member should be considered. Also, since it 
will be very difficult to provide sufficient strength in the augmented beam, the primary goal 
should be to provide adequate ductility to survive the expected rotations and continue to provide 
gravity load support and shear transfer capability. If the new concrete work is exposed to view, 
consideration should be given to the nature of the forming and surface finishes desired in relation 
to the surrounding existing elements. For cases where a deeper beam or header exists, the new 
ductile member should be designed to provide all required strength and ductility. 

Detailing Considerations 

Connection of new corbel to existing concrete elements: A typical detail showing the installation 
of a new steel corbel is shown in Figure 13.4.2-1. The steel angle provides supplemental support 
for the coupling slab at the wall joint where the most damage is expected to occur. For a precast 
plank slab condition (as shown in the detail), especially without a cast-in-place topping, 
consideration should be given to extending the supplemental steel supports along the full length 
of the link between the coupled walls. Installation of drilled anchors into the bottom of the 

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Figure 13.4.2-1: Typical Corbel at Linking Slab 

planks may be considered. However, it may be difficult to avoid both prestressing tendons and 
hollow core voids. Consider use of screen-tube anchors (see Chapter 21) if voids cannot be 
avoided. 

Connection of new concrete to the existing concrete beam and adjacent walls: Figure 13.4.2-2 
shows an elevation and section view of a typical augmentation of a relatively shallow coupling 
beam or header. The surface of the existing concrete should be thoroughly cleaned and 
roughened to provide a good bond and interaction between the existing and new portions of the 

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Figure 13.4.2-2: Typical Strengthening of Shallow Coupling Beam 

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composite member. The new concrete should be extended along the walls far enough to develop 
the longitudinal bars. If cast-in-place concrete is used, access holes will be required through the 
slab for placement and consolidation. Care must be taken to avoid damage to any prestressing 
tendons in the precast planks. Figure 13.4.2-3 shows typical details for installation of X-shaped 
rebar cages at deeper coupling beams or headers. The new concrete should be thick enough to 
allow placement of the crossing longitudinal reinforcement cages inside of the confining beam 
ties. 

Cost/Disruption Considerations 

Installation of steel corbel supplemental supports is an inexpensive, minimal mitigation 
approach. However, costs could increase substantially if relocations of M/E/P systems are 
required. The disruption associated with a single installation of this work is relatively local, but 
similar work will likely be required at all or most of the walls throughout the building. At all 
cases where new concrete or shotcrete is added to existing coupling beams, the cost and level of 
disruption will increase substantially. Refer to similar discussion in Sections 12.4.2 and 12.4.3. 

Construction Considerations 

For installation of a steel corbel, care should be taken to locate existing reinforcing steel in the 
wall prior to drilling or coring. Also, the pattern of prestressing tendons should be located before 
any drilling into precast planks is begun. For augmentation of coupling beams, refer to similar 
discussions at Sections 12.4.2 and 12.4.3. 

Proprietary Concerns 

The basic materials are generic. 

13.4.3 Enhance Connection Between Slab and Walls 
Deficiencies Addressed by the Rehabilitation Technique 

Inadequate bearing for precast slab planks at wall 

Inadequate shear transfer capacity from precast plank diaphragms to shear wall 

Description of the Rehabilitation Technique 

For most cases involving inadequate bearing support for precast floor planks at the walls, the 
typical mitigation approach will be to install steel (or concrete) ledgers secured to the wall under 
the slab-wall joint. This approach can also be used to increase joint shear capacity by including 
concrete anchors drilled up into the slab. An alternative approach to increasing joint shear 
capacity is to install diagonal dowels, or shear pins, through the joint. Occasionally, if adequate 
story height is available and gravity load issues can be addressed adequately, a new reinforced 
concrete topping slab could be placed over the existing diaphragm. 

Design Considerations 

For cases where the concern is focused only on providing adequate bearing, installation of a steel 
ledger angle anchored to the wall is a very simple and direct technique. Provision of about six 
inches of additional bearing will be enough to assure against loss of vertical support, so anchors 
drilled into the slab or precast concrete planks may be omitted. Use of a concrete ledger is 

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Figure 13.4.2-3: Strengthening of Deep Coupling Beam 

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certainly possible, but would seem appropriate only if concrete collectors were also being 
installed. However, a steel angle or channel section ledger could still be used, linking the end of 
a concrete collector to the wall. 

For cases where diaphragm-wall shear transfer is to be augmented, either vertical slab anchors or 
diagonal dowels, drilled and placed from above, may be added. However, the added shear 
capacity that can be provided by these techniques is limited by the shear capacity of the 
diaphragm slab immediately adjacent to the wall. If the local diaphragm capacity is inadequate, 
then a new collector will be required to engage a greater extent of the diaphragm (refer to 
Section 12.4.3). 

If a new concrete topping slab is to be placed, sufficient reinforcement probably can be included 
to serve as the collector. However, new diagonal dowels or perhaps a new ledger will likely be 
needed to insure transfer of the collected diaphragm shear demand down through the lightly 
reinforced slab-wall construction joints into the wall below. 

Consideration should be given to the treatment of any architectural finishes and M/E/P system 
components mounted on the wall in the affected areas. 

Detailing Considerations 

Typical details for installation of new steel ledgers and drilled diagonal dowels are shown in 
Figures 13.4.3-1 and 13.4.3-2, respectively. For these techniques, little or no prior cleaning or 
preparation of the existing concrete surfaces will be required. Alternatively, refer to Section 

12.4.3 for discussion related to installation of a concrete ledger. Where drilled concrete anchors 
or drilled diagonal dowels (acting as shear pins through the slab-wall joint region) are to be 
installed through precast concrete planks, the prestressed tendons must be avoided. Also, 
consideration should be given to means of dealing with any hollow core voids in the planks that 
may be encountered. Screen tubes similar to those used in brick masonry anchorage details could 
be used, or the voids could be filled, at least locally, with grout. 
Cost/Disruption 

Installation of steel section ledgers and drilled concrete anchors and dowels are simple and well 
known basic techniques that should be relatively inexpensive. There will be noise and vibration 
associated with the drilling, but the work is essentially dry and not particularly messy. 
However, given the nature of this type of structure, there are likely to be many walls distributed 
throughout the building, thus the work will be pervasive. If concrete ledgers are installed, the 
level of cost and disruption will be considerably higher (refer to the discussions regarding 
installation of concrete collectors at Section 12.4.3). 

Construction Considerations 

Installation of the drilled concrete anchors or diagonal dowels may require some precision to 
avoid prestressing tendons or hollow core voids. Once the pattern of these items is defined, a 
steel template, perhaps the steel ledger itself, can be used. If some flexibility in the exact location 
of the anchors is required, consider using oversized holes and welded plate washers, and provide 
extra holes. 

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Figure 13.4.3-1: Added Support and Shear Strength at Slab-Wall Joint 

The drilled anchors and dowels will require testing, which should be performed by persons 
experienced in torque and/or tension testing of diagonal dowels and overhead installations. For 
the drilled shear pin dowels that do not project out of the slab, additional similar dowels can be 
installed for testing purposes only. 

Proprietary Concerns 

The basic materials are generic. 

13.4.4 Reduce Flexural Capacity of Shear Walls to Reduce Shear Demand 
Deficiency Addressed by the Rehabilitation Technique 

Degradation of shear-critical walls 

Description of the Rehabilitation Technique 

In buildings with many walls and high strength, many of the walls may be shear-critical and 
prone to strength degradation and/or possible reduction of gravity support capacity. A sudden 

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Figure 13.4.3-2: Added Shear Capacity at Slab-Wall Joint 

loss of lateral strength and stiffness, particularly all at one floor, should be avoided. To avoid 
this condition, an alternative to shear enhancement is to reduce the level of possible shear 
demand that the wall can experience by reducing flexural capacity. This approach may be 
applied to one or more walls, in a system with adequate global strength, to change the expected 
post-yield behavior from a brittle shear failure to a more ductile flexural yielding. This approach 
also may be applied in other circumstances to an individual wall that cannot be strengthened for 
some reason, but must be protected from serious damage, even as other walls are strengthened or 
new walls are added in other locations within the system. This technique will generally involve 
cutting a number of the existing wall chord bars. However, an alternative approach that may be 
applied to a long wall would be to make one or more vertical cuts in the wall, creating a series of 
shorter panels with reduced flexural capacity. 

Design Considerations 

Application of this technique to a particular shear wall is intended to replace a brittle shear 
failure mechanism with a more ductile flexural hinging mechanism. It is not recommended that 
this technique be applied in conditions that may result in creating a brittle tension failure 
mechanism. Therefore, the number and location of cut bars must be determined with careful 
analysis of the detail of the expected mechanism (strain compatibility, flexural hinge length, 
etc.). Bar cuts should be sufficiently staggered and a major reduction in flexural capacity should 
not be attempted with this method. The impact of this technique on architectural and M/E/P 

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systems and components is likely to be relatively limited. Only fixtures in the immediate vicinity 
of the chord bar cuts or vertical wall slices are likely to be affected. However, client perception 
of this reductive technique may be decidedly negative. 

Detailing Considerations 

Cutting the selected existing wall chord bars must be done with great care and precision to avoid 
cutting adjacent chord bars or the confining transverse reinforcement. If the wall reinforcement 
can be adequately mapped by using metal detector, x-ray or ground penetrating radar techniques, 
then the selected bars could be cut by coring through the concrete cover. However, congested 
chord reinforcement and/or closely spaced transverse ties may require that the concrete cover be 
chipped away to expose the bars before cutting with either a core drill or a torch. 

In the case of reducing the overall length of a long wall with vertical cuts, the cuts can be made 
with a circular concrete saw. In order to extend the cuts as close to the floor slabs as possible the 
cuts may be made from each side of the wall, to keep the depth of the cut and the radius of the 
saw blade to about half the wall thickness. 

Cost/Disruption 

This technique will be less costly than alternative methods to increase the shear capacity of a 
wall. The extent of the work is very localized and the impact on nonstructural components will 
be limited. However, there will always be noise and vibration associated with any concrete 
chipping, coring or sawcutting, and the latter two operations can be very wet and messy. 

Construction Considerations 

Access must be available to the locations where chord bars are to be cut or the wall is to be 
sawcut. However, this will certainly be less of an imposition than any other alternative to 
increase the shear capacity of the wall. 

Employment of this technique will most likely require special scheduling and/or sequencing 
considerations relative to the other strengthening work, to avoid creating a weakened structure 
during the course of the overall retrofit project. 

Proprietary Concerns 

The basic materials are generic. 

13.5 References 
ACI 440.2R-02, 2002, Guide for the Design and Construction of Externally Bonded FRP 
Systems for Strengthening Concrete Structures, American Concrete Institute, ACI Committee 
440, Chicago, IL 

ICRI (International Concrete Repair Institute), 2003, Guide for Surface Preparation for Repair 
of Deteriorated Concrete Resulting from Reinforcing Steel Corrosion, Publication 03730, 
International Concrete Repair Institute. 

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FIB (Federation Internationale de Beton), 2001, Externally Bonded FRP Reinforcement for RC 
Structures, Technical Report, Task Group 9.3 FRP Reinforcement for Concrete Structures 
Bulletin 14, July. 

Ghobarah, A., 2004, Seismic Retrofit of RC Walls Using Fibre Composites, McMaster 
University, Hamilton, Onterio, Canada. 

Kaw, A.K., 1997, Mechanics of Composite Materials, CRC Press LLC, FL. 

Laursen, P.T., Seible, F., and G.A. Hegemier, 1995, Seismic Retrofit and Repair of Reinforced 
Concrete with Ovelays, University of California, San Diego, CA. 

Priestley, N., Seible, F. and G.M. Calvi, 1996, Seismic Design and Retrofit of Bridges, John 
Wiley & Sons, New York, NY. 

Teng, J.G., J.F. Chen, S.T. Smith, and L. Lam, 2002, FRP Strengthened RC Structures, John 
Wiley & Sons, New York, NY. 

13-24 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 14  Type C2f: Concrete Shear Walls 

Chapter 14 - Building Type C2f: Concrete Shear Walls 
(Gravity Frame Systems) 

14.1 Description of the Model Building Type 
Reinforced concrete walls in a building will act as shear walls whether designed for that purpose 
or not. Therefore, concrete buildings that contain any significant amount of concrete wall will 
fall into this category. However, there are two distinctly different types of concrete wall 
buildings: those that contain an essentially complete beam/slab and column gravity system, and 
those that use bearing walls to support gravity load and have only incidental beam and column 
framing. In this document, these building types have been separated and are designated C2f for 
the gravity frame system and C2b for the bearing wall. This section covers the building with 
gravity framing system. Although it is typically assumed that the gravity framing is not part of 
the lateral force-resisting system, the framing could add stiffness to the building, particularly 
near the top of taller buildings. This building type is very common and has been used in a wide 
variety of occupancies and in all sizes. 

Figure 14.1-1: Building Type C2f: Concrete Shear Walls 
(Gravity Frame Systems) 

14-1 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 14  Type C2f: Concrete Shear Walls 

Variations in Framing Systems 

There are wide overall variations within this building type due to the possible configuration and 
extent of the concrete walls, the many types of vertical framing systems used, and the lateral 
stiffness interaction between the two. In buildings with incidental concrete walls and a 
substantial beam-column gravity frame system, this building type merges with Building 
Type C1. If the building type is unclear, reference should be made to both Chapter 12 and this 
chapter. 

Gravity frame systems in this building type include cast-in place concrete beam and slab, one-
way joists, two-way waffles, and two-way or flat slabs. 

In older buildings that are seismically deficient, the walls were often intended for fire protection 
of vertical shafts, as exterior closure walls, or as bearing walls. However, buildings built in 
regions of high seismicity in the 1950s, 1960s or early 1970s often were designed with a shear 
wall lateral force-resisting system, but they are now found deficient due to low global strength, a 
highly torsional plan layout or detailing that leads to premature shear failure 

In buildings designed with shear walls, the walls are either strategically placed around the plan, 
or at the perimeter. Shear walls systems placed around the entire perimeter almost always 
contain windows and other perimeter openings and are often called punched shear walls. Older 
buildings will have concrete walls somewhat arbitrarily placed in plan. 

Floor and Roof Diaphragms 

The floor and roof diaphragms in this building type are essentially the same as the bearing wall 
system and are almost always cast-in-place concrete. The diaphragms are stiff and strong in 
shear because the horizontal slab portion of the gravity system is either thick or frequently 
braced with joists. However, one way joist systems could be inadequate in shear in the direction 
parallel to the joists. Collectors are seldom in place and transfer of load from diaphragm to shear 
wall must be carefully considered. 

Foundations 

There is no typical foundation for this building category. Foundations could be found of every 
type depending on the height of the building, the span of the gravity system and the site soil. 

14.2 Seismic Response Characteristics 
Shear wall buildings, unless configured with only incidental or minimal walls, will typically be 
quite stiff. Elastic and early post-elastic response will therefore be characterized with lower-than 
average drifts and higher-than-average floor accelerations. Damage in this range of response 
should be minimal. 

Overall post-elastic response is highly dependent on the specific characteristics of the shear walls 
and the gravity frame components. 

14-2 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 14  Type C2f: Concrete Shear Walls 

Shear Wall Behavior 

The walls must first be evaluated to determine if they contribute sufficient strength or stiffness to 
be considered significant. Often walls around vertical shafts are thin and lightly reinforced and 
will have little effect on the overall building response. Although retrofit techniques are similar 
such a building may be classified as Type C1. 

When subjected to ever increasing lateral load, individual shear walls or piers will first often 
force yielding in spandrels or other horizontal components restricting their drift, and eventually 
either rock on their foundations, suffer shear cracking and yielding, or form a flexural hinge near 
the base. Shear and flexural behavior is quite different, and estimates of the controlling action 
are affected by the distribution of lateral loads over the height of the structure. 

Yielding of spandrels or other coupling beams can cause a significant loss of stiffness in the 
structure. Flexural yielding will tend to maintain the strength of the system, but shear yielding, 
unless well detailed, will degrade the strength of the coupling component and the individual 
shear wall or pier will begin to act as a cantilever from its base. 

Rocking is often beneficial, limiting the response of the superstructure. However, the amplified 
drift in the superstructure from rocking must be considered. In addition, if varying wall lengths 
or different foundation conditions lead to isolated or sequencing rocking, the transfer of load 
from rocking walls must be investigated. In buildings with basements, the couple created from 
horizontal restraint at the ground floor diaphragm and the basement floor/foundation (often 
termed the backstay effect) may be stiffer and stronger than the rocking restraint at the 
foundation and should be considered in those configurations. 

Shear cracking and yielding of the wall itself is generally considered undesirable, because the 
strength and stiffness will quickly degrade, increasing drifts in general, as well as potentially 
creating a soft story or torsional response. However, in accordance with FEMA 356 (FEMA 
2000), shear yielding walls or systems can be shown to be adequate for small target 
displacements. 

Flexural hinging is considered ductile in FEMA 356 and will degrade the strength of the wall 
only for larger drifts. Similar to rocking, the global effect of the loss of stiffness of a hinging 
wall must be investigated. 

Gravity Frame Behavior 

The lateral strength and stiffness of gravity frames will vary considerably among buildings in 
this type. In some configurations of this building type, the gravity frame will not significantly 
participate in the response. However, it is not uncommon in these buildings for a stiff and brittle 
gravity system to dominate both response and the extent of damage. For example, if concrete 
spandrels or sills on the perimeter of the building restrain the gravity columns (the short 
column), the column must take the full story drift over a short height, potentially causing shear 
failure and loss of gravity load capacity. Other gravity systems, such as flat slab or heavy beam 
and column systems, can also be sensitive to drifts, particularly to the increased drifts near the 
top of buildings with walls of a height-to-width ratio over 3. The frame action of the gravity 

14-3 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 14  Type C2f: Concrete Shear Walls 

system of these buildings may be beneficial or could form a deficiency, but in any case the 
interaction with the shear walls should be considered. 

14.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
See Table 14.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 
Selected deficiencies are further discussed below by category. 

Global Strength 

Older buildings placed in this category may have minimal shear walls and seismic displacements 
will likely put excessive demand on the walls regardless of the shear or flexural behavior. The 
most common method of mitigating this deficiency is to add more shear walls, although for 
smaller buildings steel braced frames have been used. The nonlinear behavior of steel braced 
frames must be carefully studied for compatibility with the other shear walls and the gravity 
frame. Additional bending resistance of existing walls can be obtained by enhancing existing 
chords, although such walls should not be strengthened to become shear critical. If the majority 
of walls are shear-critical and strength degradation is the primary concern, shear strength can be 
added to the existing walls with concrete or FRP overlays. 

Global Stiffness 

The most important issue in a building of this type that exhibits large interstory drifts is the 
ability of the gravity system to accept the drifts while sustaining their loads. Excess drifts could 
be caused by inadequate length of wall, by rocking at the foundations, or, at the upper stories, by 
a deformed shape characterized by bending deformations. In most buildings, strength and 
stiffness are closely related, and inadequate stiffness is mitigated by adding new elements or 
stiffening existing ones, which generally will also increase strength. Damping can also be added 
to reduce drifts but care must be taken to achieve the desired damping with small displacement 
expected in a shear wall building. 

Configuration 

The two most common configuration deficiencies in this building type are 1) severe torsion 
caused by eccentrically placed shafts or towers, and 2) shear walls or full stories that are weak or 
soft from openings in the walls or from discontinuous walls that may not run through the ground 
floor or basement floor. Completely discontinuous walls also create a load transfer deficiency 
for both overturning and shear. 

Load Path 

Common load path deficiencies include discontinuous shear walls, as discussed above, and 
collectors for the shear walls. Collectors can be added with steel members or new reinforcing 
and concrete. 

Component Detailing 

Shear walls in most older buildings meet none of the current detailing requirements covering 
minimum shear reinforcement, for confinement of chords, and the walls are commonly shear 
critical. FEMA 356 allows these deficiencies at controlled displacement levels. A common 

14-4 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 14  Type C2f: Concrete Shear Walls 

Table 14.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2f Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Global 
Strength 
Insufficient in-
plane wall shear 
strength 
Concrete/masonry 
shear wall [12.4.2] 
Steel braced frame 
[12.4.1] 
Steel plate shear 
wall 
Concrete wall 
overlay [21.4.5] 
Fiber composite wall 
overlay [13.4.1] 
Steel wall overlay 
Seismic isolation 
[24.3] 
Reduce flexural 
capacity [13.4.4] 
Insufficient 
flexural capacity 
Concrete/masonry 
shear wall [12.4.2] 
Steel braced frame 
[12.4.1] 
Add or enhance 
chords [12.4.3] 
Inadequate 
capacity of 
coupling beams 
Concrete/masonry 
shear wall [12.4.2] 
Steel Braced frame 
[12.4.1] 
Strengthen beams 
Improve ductility of 
beams [13.4.2] 
Remove beams 
Global 
Stiffness 
Excess drift 
(normally near the 
top of the 
building) 
Concrete/masonry 
shear wall [12.4.2] 
Steel braced frame 
[12.4.1] 
Fiber composite 
column wrap 
[12.4.4] 
Concrete/steel 
column jacket 
[12.4.5] 
Provide detailing of 
all other elements to 
accept drifts 
Thicken walls 
Supplemental 
damping [24.4] 
Configuration Discontinuous 
walls 
Concrete/masonry 
shear wall [12.4.2] 
Enhance existing 
column for 
overturning loads 
Improve connection 
to diaphragm 
[13.4.3] 
Remove wall 
Soft story or weak 
story 
Add strength or 
stiffness in story to 
match balance of 
floors 

14-5 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 14  Type C2f: Concrete Shear Walls 

Table 14.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2f Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Configuration 
(continued) 
Re-entrant corner Add floor area to 
minimize effect of 
corner 
Provide chords in 
diaphragm [12.4.3] 
Torsional layout Add balancing 
walls, braced 
frames, or moment 
frames 
Load Path Inadequate 
collector 
Add steel collector 
[12.4.3] 
Add concrete 
collector [12.4.3] 
Strengthen existing 
beam or slab 
Enhance splices or 
connections of 
existing beams 
Discontinuous 
Walls 
Provide new wall 
support components 
to resist the 
maximum expected 
overturning moment 
Strengthen the 
existing support 
columns for the 
maximum expected 
overturning moment 
Provide elements to 
distribute the shear 
into the diaphragm at 
the level of 
discontinuity 
Component 
Detailing 
Wall inadequate 
for out-of-plane 
bending 
Add strongbacks Concrete wall 
overlay [21.4.8] 
Wall shear critical Concrete wall 
overlay [21.4.8] 
Fiber composite wall 
overlay [13.4.1] 
Reduce flexural 
capacity of wall 
[13.4.4] 
Inadequate 
displacement 
capacity of gravity 
columns 
Enhance ductility 
(see also global 
stiffness) 

14-6 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 14  Type C2f: Concrete Shear Walls 

Table 14.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2f Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Diaphragms Inadequate in-
plane shear 
capacity 
Concrete topping 
slab overlay 
Fiber composite 
overlays [22.2.5] 
Inadequate chord 
capacity 
New concrete or 
steel chord member 
[12.4.3] 
Excessive stresses 
at openings and 
irregularities 
Add chords [12.4.3] Infill openings 
[22.2.4] 
Foundations See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 

14-7 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 14  Type C2f: Concrete Shear Walls 

improvement to these walls is to enhance shear strength to be equal or greater than the maximum 
that can be developed in the wall, based on bending strength. 

Diaphragm Deficiencies 

The most common diaphragm deficiency in this building type is a lack of adequate collectors. 
The addition of effective collectors in an existing diaphragm is difficult and disruptive. Existing 
strength to deliver loads to the shear walls should be studied carefully before adding new 
collectors. 

Foundation Deficiencies 

See Chapter 23. 

14.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
The most relevant recommendations listed in Table 14.3-1 are similar to techniques also 
associated with other concrete building types such as C1 and C2b or general techniques applied 
to concrete diaphragms. Details concerning these techniques can be found in other chapters. 

14.5 References 
FEMA, 2000, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 
356, Federal Emergency Management Agency, Washington, D.C. 

14-8 


Techniques for the Seismic Rehabilitation of Chapter 15Types C3/C3A Concrete Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls 

Chapter 15 - Building Types C3/C3A: Concrete Frames with 
Infill Masonry Shear Walls 

15.1 Description of the Model Building Type 
Building Type C3 is normally an older building that consists of an essentially complete gravity 
frame assembly of concrete columns and floor systems. The floors consist of a variety of 
concrete systems including flat plates, two-way slabs, and beam and slab. Exterior walls, and 
possibly some interior walls, are constructed of unreinforced masonry, tightly infilling the space 
between columns horizontally and between floor structural elements vertically, such that the 
infill interacts with the frame to form a lateral force-resisting element. Windows and doors may 
be present in the infill walls. The buildings intended to fall into this category often feature 
exposed clay brick masonry on the exterior. Figure 15.1-1 shows an example of this building 
type. 

It is important to note that similar buildings with exterior masonry infill sills below windows that 
extend column to column do not behave with strut action and should be classified as Building 
Type C1 or C2f. In fact, such infill sills often create short columns that must absorb the entire 
story drift over the unrestrained height of the window, which can often be an extreme deficiency 
in poorly reinforced columns. The C3A building type is similar but has floors and roof that act 
as flexible diaphragms such as wood, or untopped metal deck. 

Figure 15.1-1: Building Type C3: Concrete Frames with Infill Masonry Shear Walls 

15-1 


Techniques for the Seismic Rehabilitation of Chapter 15Types C3/C3A Concrete Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

Variations Within the Building Type 

The building type was identified primarily to capture the issues of interaction between 
unreinforced masonry and concrete gravity framing. The archetypical building has solid clay 
brick at the exterior with one wythe of brick running continuously past the plane of the column 
and beam and two or more wythes infilled within the plane of the column and beam. The 
exterior wythe of clay brick forms the finish of the building although patterns of terra cotta, 
stone, or precast concrete may be embedded into the brick. However, there can be many 
variations to this pattern depending on the number and arrangement of finished planes on the 
exterior of the building. For example, the full width of the infill wall may be located with the 
plane of the column and beam with a pilaster built out and around the column and a horizontal 
band of brick or other material covering the beam; the beam may also be slightly offset from the 
centerline of the column to accommodate the pattern of exterior finishes. 

Hollow clay tile masonry may also be used as an exterior infill material. Although this material 
often has a high compression strength, the net section of material available to form the 
compression strut within the frame will normally contribute a lateral strength of only a small 
percentage of the building weight. The material being brittle and the wall being highly voided, 
these walls may also lose complete compressive strength quite suddenly. Therefore, walls of 
hollow clay tile infill will probably not contribute a significant portion of required lateral 
resistance except in areas of low seismicity and/or when walls are arranged as infill on both the 
exterior and interior of the building. 

More recent buildings may have unreinforced concrete block masonry configured as an exterior 
infill wall, with a variety of finish materials attached to the outside face of the concrete block. 
Similar to hollow clay tile walls, these walls may exhibit moderate to low compressive strength 
and brittle behavior that marginalizes their usefulness as lateral elements. In addition, hollow 
concrete block exterior walls often will not be installed tight to the surrounding framing, 
eliminating infill compression strut behavior. 

Floor and Roof Diaphragms 

Floors are often flat plates or two-way slabs. Beam and slab or beam and joist systems will also 
be found in this building type. Typically, these slabs provide adequate diaphragms. 

Building Type C3A will have heavy timber floors with one or more layers of sheathing forming 
a diaphragm. The flexibility of such diaphragms will often form a seismic deficiency because, 
assuming no interior shear elements, the large drift at the diaphragm mid-span will damage 
perpendicular walls and gravity framing. Specific strengthening techniques for this building type 
are not covered here. For generalized strengthening of diaphragms, see Chapter 22. 

Foundations 

There is no typical foundation for this building type. Foundations can be found of every type 
depending on the height of the building, the span of the gravity system and the site soil. The 
exterior walls are exceptionally heavy and typically will be supported by a continuous concrete 
footing or often a continuous concrete wall forming a basement space below. 

15-2 


Techniques for the Seismic Rehabilitation of Chapter 15Types C3/C3A Concrete Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

15.2 Seismic Response Characteristics 
Both in terms of stiffness and strength, the exterior infill walls will form the effective lateral 
system for this building type. The effectiveness of the system depends on the size and extent of 
openings and articulation of the plane of the wall. With solid or nearly solid infill panels, strut 
action will be stiff and strong. As openings in panels increase in size, struts or combinations of 
struts cannot effectively form around the opening, and the concrete columns and beams may 
begin to work as a moment frame, with fixity at the beam-column joint provided by the 
masonry. For low and moderate intensity shaking, the exterior walls may provide adequate 
strength to satisfy the specified performance objective. As the shaking demand increases, the 
masonry will tend to crack and spall, losing stiffness and potentially creating a falling hazard. 
The complete concrete gravity system, characteristic of this building type, will provide additional 
stability, but may quickly degrade in strength and stiffness due to inadequate column reinforcing. 
However, in wall configurations with large height-to-width ratios, end or corner columns could 
experience large compression or tension loads from overturning, leading to rapid degradation of 
column lateral and gravity capacity. 

This building type is often characterized by a commercial store-front first floor with little or no 
infill at that level on one or more faces of the building. This condition can cause a soft story 
condition or a severe torsional response if open on one or two sides only. Such conditions can 
lead to concentration of seismic deformation at the open level, degradation of the columns, and 
possible P-delta failure. 

15.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
See Table 15.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 
Deficiencies related to specifically to concrete moment frames and to masonry shear walls are 
shown in Table 12.3-1 and Table 18.3-1, respectively. Selected deficiencies are further 
discussed below by category. 

Global Strength 

The overall strength provided by the exterior walls may be insufficient to prevent serious 
degradation and resulting amplified displacements that can lead to irreparable damage or even 
instability. The strength may be limited by inadequate number of panels of infill, excessive 
openings, or masonry weak in compressive strength. The standard approach to such deficiencies 
will be to add new, relatively stiff lateral force-resisting elements such as concrete shear walls or 
steel braced frames often located on the interior between existing columns. The infill itself could 
also be strengthened by adding a layer of reinforced concrete on the interior surface, although 
whether such a system is acting as infill or as a shear wall must be checked by analysis. Unless 
all infill has been backed by a support system, the damage state of the infill wall must be 
estimated for the expected drifts of the combined system to determine if the desired performance 
has been achieved. 

Global Stiffness 

The most important issue in a building of this type that exhibits large interstory drifts is the 
ability of the gravity system to accept the drifts while sustaining their loads. In most buildings, 

15-3 


Techniques for the Seismic Rehabilitation of Chapter 15Types C3/C3A Concrete Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

Table 15.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C3/C3A Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Global 
Strength 
Inadequate length 
of exterior wall 
Interior concrete 
walls [12.4.2] 
Interior steel braced 
frames [12.4.1] 
Concrete wall overlay 
[21.4.5] 
Fiber composite wall 
overlay [21.4.6] 
Excessive sized 
openings in infill 
panels 
Interior concrete 
walls [12.4.2] 
Interior steel braced 
frames [12.4.1] 
Infill selected 
openings [21.4.7] 
Fiber composite wall 
overlay [21.4.6] 
Inadequate 
columns for 
overturning forces 
Add confinement 
Add tensile capacity 
on outside surface of 
column. 
Weak or 
deteriorated 
masonry 
Interior concrete 
walls [12.4.2] 
Interior steel braced 
frames [12.4.1] 
Point outside and/or 
inside wythes of 
masonry 
Inject wall with 
cementitious grout 
Fiber composite 
overlay [21.4.6] 
Global 
Stiffness 
See Global 
Strength 
Configuration Soft or weak story Interior concrete 
walls [12.4.2] 
Interior steel braced 
frames [12.4.1] 
Torsion from one 
or more solid 
walls 
Balance with 
Interior concrete 
walls 
Balance interior 
steel braced frames 
Remove selected 
infill panels on 
solid walls 

15-4 


Techniques for the Seismic Rehabilitation of Chapter 15Types C3/C3A Concrete Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

Table 15.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C3/C3A Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Configuration 
(continued) 
Irregular Plan 
Shape 
Balance with 
Interior concrete 
walls 
Balance with 
Interior steel braced 
frames 
Load Path Out-of-plane 
failure of infill 
due to loss of 
anchorage or 
slenderness of 
infill 
Provide surface wall 
supports [13.4.3] 
Shotcrete overlays 
[21.4.5] 
Fiber composite 
overlay [21.4.6] 
Remove infill 
Inadequate 
connection of 
finish wythe to 
backing 
Add connections 
Inadequate 
collectors 
Add steel collector 
or concrete collector 
[12.4.3] 
Component 
Detailing 
Inadequate 
columns splice for 
tension due to 
uplift force 
induced by infill 
Add splice plates 
Provide splice 
through added 
reinforced concrete 
encasement 
Inadequate beam 
column 
connection to 
resist compression 
thrust 
Strengthen connection 
in shear with steel or 
fiber composite 

15-5 


Techniques for the Seismic Rehabilitation of Chapter 15Types C3/C3A Concrete Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

Table 15.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C3/C3A Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Component 
Detailing 
(continued) 
Weak or 
incompletely 
filled joint 
between masonry 
and surrounding 
steel components 
Create appropriate 
clean void and repack 
with masonry and/or 
mortar. 
Inject voids with 
cementitious grout or 
epoxy 
Diaphragms See Chapter 22 
Foundations See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 

15-6 


Techniques for the Seismic Rehabilitation of Chapter 15Types C3/C3A Concrete Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls 

strength and stiffness are closely related, and inadequate stiffness is mitigated by adding new 
elements or stiffening existing ones, which generally will also increase strength. Damping can 
also be added to reduce drifts, but care must be taken to achieve the desired damping with the 
small displacement expected in a shear wall building. 

Configuration 

Two configurational deficiencies are common in this building type. The first is a soft and weak 
story at the street level created by commercial occupancies with exterior bays with little or no 
infill. This deficiency can be corrected by adding selected bays of infill or by adding shear walls 
or braced frames at this level. The second common issue is a plan torsional irregularity created 
by solid masonry walls on property lines coupled with walls with many openings on street fronts. 
If shown by analysis to be necessary, torsional response can be minimized by stiffening the more 
flexible side of the building with more infill or by the addition of lateral elements. In rare cases, 
the solid walls can be balanced with the open side by selected removal of panels or 
disengagement of the infill strut action. 

Load Path 

The primary load path issue with this building type is to assure that the mass of the exterior walls 
will not become disengaged from the frame which will both prevent infill strut action as well as 
to create a significant falling hazard on the street below. 

In-plane, the articulation of the exterior walls may result in offsets of the wall plane between 
floors. The presence of a complete load path and maintenance of confinement for strut formation 
must be reviewed in such instances. 

If new lateral load-resisting elements are added, existing slab and/or beam construction may need 
to be strengthened to provide adequate collectors. 

Component Detailing 

In order to qualify as an infill lateral force-resisting element, the infill must be installed tight to 
the surrounding concrete elements. Loose or incomplete infill can be mitigated with local 
patching of the masonry or by injection of cemetitious or epoxy grout. 

As previously noted, the infill adjacent to columns must be of sufficient stiffness to provide a 
floor to floor diagonal strut. With narrow piers surrounding columns, the jamb and header 
masonry may restrain the column such that the entire story drift must be absorbed in the central 
length of the column, often creating dangerous shear failures. 

The detailing of the concrete frame forming the confinement for the masonry is important to 
achieve infill strut behavior. The connection of beam to column must be capable of resisting the 
strut compression forces from the masonry. The shear capacity of the beam-to-column 
connection is often critical. Strengthening of these connections may require removal of 
considerable masonry to obtain adequate access. Some of the techniques developed to 
strengthen concrete moment frames may be applicable. In addition, splices in vertical 
reinforcing of column splices may be inadequate to transfer the tensile overturning forces created 

15-7 


Techniques for the Seismic Rehabilitation of Chapter 15Types C3/C3A Concrete Frames 
Existing Buildings: FEMA 547 with Infill Masonry Shear Walls 

by strut action. These areas can be confined to reduce the required splice length or augmented 
with additional tensile strength from the surface. 

Diaphragm Deficiencies 

Concrete slab diaphragms often will be adequate. The connection of slab to exterior wall should 
be reviewed. 

See Chapter 21 on URM construction for discussion of wood diaphragms in this type of building. 

Foundation Deficiencies 

No systematic deficiency in foundations should be expected solely due to the characteristics of 
this building type. 

15.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
Most significant recommendations listed in Table 15.3-1 are similar to techniques more 
commonly associated with other building types such as various concrete buildings (C1, C2b, or 
C2f), unreinforced masonry bearing wall buildings (URM), or general techniques applied to 
concrete diaphragms. Details concerning these techniques can be found in other chapters. 

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Techniques for the Seismic Rehabilitation of 
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Chapter 16 - Building Type PC1: Tilt-up Concrete Shear Walls 

16.1 Description of the Model Building Type 
Building Type PC1 is constructed with concrete walls, cast on site and tilted up to form the 
exterior of the building. PC1 buildings are used for many occupancy types including warehouse, 
light industrial, wholesale and retail stores, and office. The majority of these buildings are one 
story; however, tilt-up buildings of up to three and four stories are common, and a limited 
number with more stories exist. For many years, tilt-up buildings have been primarily large box-
type buildings with the tilt-up walls at the building perimeter; this is by far the largest group of 
PC1 buildings in the U.S. inventory. In recent years, tilt-up construction has expanded to more 
varied uses and building configurations. Figure 16.1-1 illustrates one example of a PC1 building. 

Figure 16.1-1: Building Type PC1: Tilt-Up Concrete Shear Walls 

Key to the PC1 building type as addressed by this chapter is the combination of a flexible roof 
diaphragm and rigid walls. Lateral forces in PC1 buildings are resisted by flexible wood 
sheathed or steel deck roof diaphragms, wood, composite steel deck, or precast floor diaphragms, 
and tilt-up concrete shear walls. In some local areas, walls are concrete panels or T-beams cast 
off-site, rather than on-site; PC1 also applies to this variation. 

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One-story PC1 buildings with flexible roof diaphragms are the primary focus of this chapter; 
however, discussion of wall-to-flexible diaphragm anchorage is equally applicable to the roof of 
multistory PC1 buildings. For other rehabilitation issues in multistory buildings, refer to the C1 
Building Type. For variations with rigid diaphragms at floors and roof, see Building Type PC2. 

Guidelines for Seismic Evaluation and Rehabilitation of Tilt-Up Buildings and Other Rigid 
Wall/Flexible Diaphragm Structures or SEAONC Guidelines (SEAONC, 2001) provides a 
substantial collection of information on West Coast PC1 building configurations, experience 
with earthquake performance, rehabilitation priorities, and techniques for rehabilitation. This 
chapter highlights the major rehabilitation considerations from the SEAONC Guidelines 
document, and provides suggested adaptations for construction variations. Provisions addressing 
rehabilitation can also be found in International Existing Building Code (IEBC) Appendix 
Chapter A2 (ICC, 2003b), City of Los Angeles Building Code Chapter 91 (City of Los Angeles, 
2002) and Guidelines for Seismic Retrofit of Existing Buildings (GSREB) (ICBO, 2001). These 
provisions focus on wall anchors, diaphragm cross-ties, and collectors, with the goal of hazard 
reduction. This chapter and the SEAONC Guidelines address a broader range of rehabilitation 
issues and techniques. 

The Tilt-up Construction and Engineering Manual (Tilt-Up Concrete Association, 2004), now in 
its sixth edition, serves as a primary resource for design and detailing practice for new tilt-up 
construction. 

Walls 

Tilt-up exterior walls are the primary vertical elements in the lateral force-resisting system. Tilt-
up buildings with large plan areas may have interior tilt-up walls or braced steel frames 
providing additional lateral resistance. Large-scale construction of tilt-up buildings began in the 
1950s and 1960s with primarily solid wall panels for warehouse and light-industrial use. As tilt-
up construction expanded to commercial use in the 1970s and 1980s, the wall panels changed to 
include large window and door openings and multistory construction. The publications 
Recommended Tilt-up Wall Design (SEAOSC, 1979) and Test Report on Slender Walls (ACISEASC, 
1982) document the change from use of code-prescribed height to thickness (h/t) limits 
of 25 for bearing walls and 36 for other walls to use of much higher h/t ratios, in combination 
with rational analysis of slenderness effects. Other code changes of interest have occurred for 
wall pier reinforcing requirements and wall panel connection to the foundation or slab-on-grade. 

Most tilt-up panels are a single piece from the foundation to the top of the building; however, 
some tilt-up systems use separate wall panels at upper stories, or lintel (spandrel) panels that are 
supported on other tilt-up wall panels. Welded connections of these panels can be damaged when 
they restrict panel movement under earthquake load. Prior to start of rehabilitation design, it is 
important to identify the tilt-up panel joint and support locations and connection condition. 
Surface treatments (such as exposed aggregate) and reveal joints are commonly used to visually 
enhance tilt-up wall panels. These treatments effect the location and dimensions of critical 
sections for design. 

Connections between adjacent tilt-up panels are often relied on to provide continuity for 
diaphragm chords and collectors. Connection types have varied over the years. Early tilt-up walls 

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Techniques for the Seismic Rehabilitation of 
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were often joined by cast-in-place pilasters. Later, welded connections between cast-in 
embedments or between horizontal reinforcing steel were commonly used. Both of these 
connection systems experience some problems with fractures at welds due to panel shrinkage 
and temperature movement. As a result, the codes placed carbon equivalent requirements on 
rebar to be welded, and detailing to minimize restraint of panel movement was pursued. In 
Recommended Tilt-up Wall Design (SEAOSC, 1979), minimal interconnection between the 
panels and special detailing of chord and collector connections are suggested to allow movement. 
Detailing suggestions included 1) breaking the bond between the chord reinforcing and concrete 
for one fourth the panel width in from each panel end and 2) use of steel angle chord/collector 
members connected in the center portion of the panel, but slotted near the panel ends. The panel 
connection detailing should be considered in evaluating the behavior of the chords and collectors 
and in the distribution of shear to the tilt-up panels. 

Gravity Load Support at the Building Perimeter 

It is most common for roof and floor framing members to be supported on the tilt-up panels at 
the building perimeter; however, some contractors have found it advantageous to provide steel 
gravity columns at the inside face of the tilt-up walls. This separation of the gravity and lateral 
systems makes construction tolerances for items embedded in the tilt-up panels and construction 
sequence less critical. The use of columns requires some modification of wall anchorage details 
at girders, but has little or no effect on the typical wall to diaphragm attachments, or fundamental 
building behavior. 

Floor and Roof Diaphragms 

The PC1 roof system will generally be of either wood or steel construction. In the western states 
(primarily California, Oregon, Washington), roof systems are almost exclusively wood structural 
panel sheathed. Subpurlins (joists), purlins, and girders are most often wood; however, open web 
trusses with wood chords or nailers are also used. Wood girders are most often supported on 
steel columns at the building interior. Sheathing fastening and therefore unit shear capacity in 
wood structural panel diaphragms generally varies by nailing zone or area. In older West Coast 
PC1 buildings, roofs can be lumber sheathed, and roof framing can include bow-string trusses. 

Outside of the western states, roof systems are almost exclusively sheathed with steel decking 
with rigid foam or nonstructural concrete insulation. The roof framing system is most commonly 
of steel open web trusses (bar joists) used in combination with truss girders or hot-rolled steel 
beams. 

Interior Additions 

Mezzanines and interior second stories constructed within tilt-up buildings are common. The 
interior addition may be seismically separated and braced independently of the building shell 
(exterior walls and roof), or may be attached to and supported by the shell. Attachment to the 
building shell raises two potential issues for the tilt-up: 1) whether the seismic load to the tilt-up 
system is significantly increased beyond that considered in initial design, and 2) whether the 
interior addition restrains seismic deflections of the building and creates an unintended load path. 
Both of these issues should be addressed in evaluation and rehabilitation. 

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Foundations 

Tilt-up buildings are often constructed on continuous or isolated spread footings; however, 
drilled pier and grade beam foundations are common in some regions. Often tilt-up wall panels 
are set on top of shims and grout pads on top of the foundation. Where continuous foundations 
are provided, grout may then be placed under the full length of the panel to provide continuous 
bearing. Often, no direct connection is provided between the wall panel and foundation. While 
not common practice in older tilt-up construction, in many buildings constructed in the 1980s 
and later, wall panel connection to the slab-on-grade were provided, allowing transfer of 
horizontal seismic forces to the slab. The most common connection uses rebar dowels, cast into 
the wall panel and a slab closure strip. Other approaches include welded or bolted connections 
between cast-in connection plates or threaded inserts. See Section 16.4.5 for further discussion. 

Due to the stiffness of the wall system, tilt-up buildings are best located on sites with very stable 
soils; however, they are often relegated to poor soil sites. Where tilt-up buildings are located on 
sites with soils subject to expansion, consolidation, or liquefaction, the effects of any damage to 
the foundation and wall connections due to soil movement should be considered in building 
evaluation. 

16.2 Seismic Response Characteristics 
The seismic response of the classic large box-type one-story PC1 building is characterized by 
rigid wall and flexible diaphragm behavior. In this type of building, the concrete walls will have 
a very short period, while the diaphragm has long-period behavior. Amplification of seismic 
forces near the center of wood sheathed diaphragms loaded in the transverse direction can be 
significant. This creates high demand on the diaphragm and out-of-plane wall anchorage near the 
center of the diaphragm, and it can also generate very high shear demand at the diaphragm 
connection to the end walls. This behavior has been replicated in instrumented buildings and in 
laboratory testing (Fonseca, Wood and Hawkins, 1996). Roof diaphragm amplification in the 
longitudinal direction can be lower, depending on the diaphragm aspect ratio; however, 
significant overstrength can result in large anchorage forces in this direction also. 

Observation of the earthquake performance of these buildings has identified as important 1) 
understanding the magnitude of the wall anchorage force and 2) detailing to eliminate weak links 
in the wall anchorage connection. With each subsequent earthquake since 1971, building code 
requirements have been revised to reflect changing understanding of the magnitude of forces 
generated in wall anchors and requirements for proper detailing; simply providing positive 
connections from the walls to the diaphragm has not resulted in adequate performance. A variety 
of weaknesses in the connections have kept anchorages from developing adequate capacity. 
Included are cross-grain tension of wood ledgers, rotation of non-symmetric connectors, low-
cycle fatigue of straps that buckle under compression loads, net section fracture of straps with 
punches bolt holes, etc. The current code wall anchorage force and detailing requirements reflect 
a history of knowledge gained for wood structural panel diaphragms in large box-type buildings. 

Recent construction of PC1 buildings has moved away from this classic large box-type building 
with wood structural diaphragms. In most regions of the United States, PC1 buildings are now 
constructed with steel deck diaphragms. Many PC1 box-type buildings are being constructed 
with very tall walls, in which out-of-plane behavior could potentially modify building seismic 

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Techniques for the Seismic Rehabilitation of 
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response. Tilt-up construction has expanded to include a wide variety of building uses with 
different building characteristics, including smaller and less regular plans, less regular diaphragm 
configurations, and some with interior as well as exterior tilt-up walls. The seismic response 
characteristics of these building configurations are not known. Their behavior is of much 
interest, however, because current building code provisions may not adequately address these 
variations and materials. The Tilt-Up Concrete Association has just initiated the TCA Seismic 
Performance Initiative, with the intent of identifying and developing strategic plans to resolve 
issues of building performance and code requirements for design and detailing. Initial work will 
focus on design models developed from instrumented building behavior (Freeman, Searer and 
Gilmartin, 2002). Research on tilt-up building performance is also in progress at Canterbury 
University, New Zealand. 

16.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
PC1 buildings with wood sheathed roof diaphragms have experienced structural damage and 
partial building collapse in a number of California earthquakes, as well as the 1964 Alaska 
earthquake. Partial collapse is almost exclusively associated with inadequate connection of the 
walls to the flexible roof diaphragm for out-of-plane loading. A variety of other types of damage 
to the wall panels, connections and roof diaphragms have been observed, including some interior 
diaphragm failures observed in the Northridge earthquake. These observations have been almost 
exclusively of buildings with wood diaphragms. 

Currently used design provisions for new buildings [IBC (ICC, 2003a), ASCE 7 (ASCE, 2005), 
and NEHRP (FEMA, 2004)] contain one set of requirements addressing wall anchors and cross-
ties for concrete and masonry walls used with flexible diaphragms. These requirements do not do 
not differentiate between wood and steel deck diaphragms in application these requirements, nor 
do they differentiate between different possible tilt-up building configurations and behaviors. 
Opportunities to observe earthquake performance of steel deck diaphragms and newer building 
configurations have been limited to date. Until performance information is available from 
research or earthquake experience, the vulnerable behavior seen in wood diaphragm PC1 
buildings needs to be considered a possibility for all PC1 buildings. The fundamental 
rehabilitation concept is positive anchorage of tilt-up walls to supporting diaphragms, with 
anchorage load paths adequate for forces both away from and towards the diaphragm. 

The SEAONC Guidelines document provides a detailed list of rehabilitation measures and 
relative priorities based on both potential hazard and the level of design of the existing 
construction. The major categories of deficiencies and rehabilitation (not prioritized) are: 

Out-of-plane wall anchors to walls and pilasters 
Diaphragm cross-ties 
Collectors 
Diaphragm strength, stiffness and openings 
Wall in-plane shear connections
Wall in-plane capacity 
Wall in-plane base anchorage 
Wall out-of-plane bending 

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These deficiencies and rehabilitation measures are included in Table 16.3-1 and the general 
discussion below; however, the compilation of information in the SEAONC Guidelines is 
recommended as a useful resource. 

Global Strength and Stiffness 

Global strength and stiffness of the tilt-up walls have not been seen as a significant source of 
damage in PC1 buildings to date. This is likely due to the length of solid wall provided in older 
buildings, minimum reinforcing ratios, and the tendency to design at low stress levels so that a 
single layer of reinforcing without special detailing can be used. Strength and stiffness 
deficiencies are most likely to occur at wall lines with significant number of large penetrations 
and other locations where loads are carried by a limited number of panels. Rehabilitation 
measures include addition of new vertical elements, enhancing existing walls, and infilling 
openings in existing walls. 

Configuration 

Poor distribution of shear walls can result in torsionally irregular behavior of PC1 buildings. 
Common occurrences include an entire line of highly perforated tilt-up panels such as at loading 
dock walls in distribution and storage facilities and street front walls in commercial buildings. 
Concrete cracking and spalling have been seen in perforated wall panels that act as frames. The 
most direct approach to rehabilitation of this condition is the addition of strength and stiffness in 
line with the perforated wall. This can be accomplished through addition of new shear walls, 
enhancing of existing shear walls, or addition of steel braced frames. 

Re-entrant corners are reasonably common in large box-type PC1 buildings, either due to in-set 
panels (Figure 16.3-1), or an L-shaped building plan. The in-set walls create a hard spot in the 
diaphragm, restraining it from the deflection required to transmit load to the end wall. In most 
cases, the in-set walls will have to be considered shear walls supporting the diaphragm. Where 
chords and collectors have not been provided at the re-entrant corner, the diaphragm has been 
seen to pull away from the wall, damaging gravity and lateral load connections. Rehabilitation at 
re-entrant corners requires the provision of adequate chords and collectors, shear transfer to the 
in-set wall panels, and possibly the strengthening of the diaphragm, wall panels, and connections 
to the foundation. In some cases, high earthquake loads in existing elements may make it 
necessary to add new vertical elements in line with the re-entrant corner. The SEAONC 
Guidelines suggest that there may be diaphragm continuity over this interior diaphragm support, 
increasing the diaphragm reaction to the in-set wall line. Alternately, it may be possible to allow 
the wall at the re-entrant corner to rock, or to separate the wall from the diaphragm allowing the 
diaphragm to span to the exterior wall. These approaches require complex detailing, however. 

Some very large tilt-up buildings are constructed in configurations that resemble multiple 
buildings alongside each other, as seen in Figure 16.3-2. This figure illustrates a single large tilt-
up building constructed in three sections, separated by roof expansion joints. Thermal 
movements in the steel deck diaphragms make the expansion joints necessary. If each of the 
three building sections were provided a complete lateral force-resisting system and separated by 
adequate seismic joints, this building configuration would be of little concern. As shown in 
Figure 16.3-2, however, building sections are often laterally braced off of adjacent building 
sections, using shear transfer through the expansion joint. This configuration raises concerns of 

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Techniques for the Seismic Rehabilitation of 
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Table 16.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for PC1 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce 
Demand 
Remove Selected 
Components 
Global 
Strength 
Insufficient in-plane 
strength of shear walls 
or frames 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [21.4.8] 
Fiber composite wall 
overlay [13.4.1] 
Concrete wall overlay 
[21.4.5] 
Infill openings 
Global 
Stiffness 
Insufficient in-plane 
stiffness of shear walls 
or frames 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [21.4.8] 
Fiber composite wall 
overlay [13.4.1] 
Concrete wall overlay 
[21.4.5] 
Infill openings 
Configuration Torsionally irregular 
plans (highly 
perforated wall line) 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [21.4.8] 
Enhance existing 
collector [7.4.2] 
Fiber composite wall 
overlay [13.4.1] 
Concrete wall overlay 
[21.4.5] 
Infill openings 
Re-entrant corners Steel braced frame 
[7.4.1] 
Collector [7.4.2] 
Concrete/masonry 
shear wall [21.4.8] 
Enhance existing 
collector [7.4.2] 
Fiber composite wall 
overlay [13.4.1] 
Concrete wall overlay 
[21.4.5] 
Infill openings 
Incidental bracing Isolate 
component from 
lateral force-
resisting system 
Load Path Inadequate or missing 
wall-to-diaphragm tie 
for out-of-plane load  
exterior and interior 
walls 
Wall-to-diaphragm 
anchorage [16.4.1] plus 
diaphragm cross-ties 
[22.2.3] 

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Techniques for the Seismic Rehabilitation of 
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Table 16.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for PC1 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce 
Demand 
Remove Selected 
Components 
Load Path 
(continued) 
Inadequate anchorage 
to diaphragms for in-
plane forces  exterior 
and interior walls 
Wall-to-diaphragm 
shear anchors [21.4.8] 
Beam or girder 
connection to tilt-up 
wall inadequate for 
wall out-of-plane loads 
Enhance beam or girder 
connection [16.4.2] 
Inadequate connection 
at base of tilt-up panel 
Wall-to-foundation 
connections [16.4.3] 
Inadequate collectors Add collector 
[7.4.2] 
Enhance existing 
collector [7.4.2] 
Component 
Detailing 
Wall inadequate for 
out-of-plane bending 
Wall strongback or 
pilaster [21.4.3] 
Inadequate detailing of 
narrow wall piers 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [21.4.7] 
Supplement component 
to provide adequate load 
path [13.4.1], [21.4.5] 
Add backup vertical 
supports where bearing 
might be lost [21.4.11] 
Diaphragms Inadequate in-plane 
strength and/or 
stiffness 
Steel braced frame 
[7.4.1] 
Concrete or 
masonry wall 
[21.4.8] 
Enhance existing 
diaphragm [22.2.1] 
Horizontal braced frame 
Inadequate chord 
capacity 
Add chord [22.2.2] Enhance existing chord 
Excessive stresses at 
openings and 
irregularities 
Enhance diaphragm 
detailing 
Excessive stresses at 
openings and 
irregularities 
Enhance diaphragm 
detailing 
Foundations See Chapter 23 
[ ] Numbers in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 

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Techniques for the Seismic Rehabilitation of 
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Figure 16.3-1: Plan of PC1 Building with Re-Entrant Corner at In-Set Panels 

deformation compatibility between the building sections under earthquake loading. Incompatible 
deformations are likely to significantly compromise shear transfer through the expansion joint. In 
addition, common details used for the expansion joint may not perform adequately under 
expected seismic forces. Details often involve significant eccentricities. The eccentricities may 
create forces in roof framing members that were not envisioned in the framing member design. 
The most direct approach to rehabilitation of this building type is the addition of vertical 
elements such that each building section is independently braced. Alternatively, the expansion 
joint connection can be improved such that it can reliably transfer anticipated earthquake forces 
while accommodating anticipated building movement. 

Another configuration concern is the occurrence of components (such as mezzanines) that act as 
incidental bracing, creating an unintended load path. For the PC1 building seismic resisting 
system to work as intended, the diaphragm must be able to deflect, and the walls must be able to 
deflect out-of-plane to follow the diaphragm. Ideally rehabilitation of incidental bracing would 
involve isolation from the building shell. Alternately, the incidental bracing could be analyzed as 
part of the structural bracing system. 

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Techniques for the Seismic Rehabilitation of 
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Figure 16.3-2: Plan and Detail of Large PC1 Building Constructed in Three Sections 

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Techniques for the Seismic Rehabilitation of 
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Load Path 

As previously mentioned, out-of-plane anchorages between the tilt-up walls and the diaphragm 
have been the primary source of damage and focus of rehabilitation in wood diaphragm PC1 
buildings. Conceptually, the approach for both new construction and rehabilitation has been to 
create continuous diaphragm cross-ties between exterior walls on opposite sides of the building. 
Rehabilitation of wall out-of-plane anchorage is discussed in this chapter. Diaphragm cross-ties 
are discussed in Chapter 22. Wall anchorage for in-plane shear commonly uses different 
connectors than for out-of-plane loads. In-plane shear connection is discussed as part of the 
diaphragm chord discussion in Chapter 22. 

Girder gravity load connections to tilt-up walls provide diaphragm-to-wall anchorage and will 
therefore have to resist wall out-of-plane anchorage forces. These anchorage forces may or may 
not have been considered in initial connection design. Even if considered, the connection may 
not be adequate for currently required loads. Rehabilitation of these connections is often 
required. 

The addition of or enhancement of existing collectors may be required in order to transmit 
diaphragm forces to the resisting shear walls. This is particularly of concern when a limited 
number of solid panels are intended to carry a significant portion of the building shear. Although 
not as common, there is also significant concern when vertical offsets in the roof diaphragm 
result in incomplete chords or collectors. Any breaks or offsets in chords or collectors need to be 
carefully evaluated. In the 1993 Guam earthquake (EERI, 1995), a high bay portion of a forklift 
repair shop was braced off of lower bays on each side. Incomplete collectors from the high-bay 
diaphragm to low bay shear walls resulted in damage. 

The anchorage at the tilt-up panel at the wall base may also be deficient. In some older tilt-ups, 
no positive connections were made from the wall to the foundation or slab; friction was relied on 
for force transfer both in-plane and out-of plane. Most tilt-up panels in recently constructed 
buildings will have a base connection, either to the foundation or more likely to the adjacent 
slab-on-grade; however, it is possible for the connection to be inadequate. Rehabilitation most 
often involves the addition of new wall to slab connections. 

Component Detailing 

Component detailing deficiencies include inadequate out-of-plane wall capacity. This deficiency 
may occur due to increases in design seismic forces or inadequate consideration of panels with 
openings. It is seldom practical to address wall capacity by adding reinforcing and concrete 
thickness to individual wall sections, so addition of wall pilasters or strongbacks is common. 
Where pilasters are added to tilt-up walls, the pilasters stiffen the wall for out-of-plane forces, 
allowing two-way spanning of the wall and attracting high out-of-plane forces to the pilaster and 
pilaster-to-diaphragm connection. A pilaster-to-roof diaphragm anchorage must be provided to 
accommodate the concentrated wall out-of-plane force. 

Tilt-up panels with large openings generally have narrow wall panels on each side. These panels 
and the spandrel wall over the opening act as a concrete frame. In loading dock and storage 
facilities, entire walls can be made up of frames. Because the narrow wall piers do not meet the 
ACI 318 (ACI, 2005) definition of a column, many have been constructed with standard wall 

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detailing. The 1994 and later editions on the UBC (ICBO, 1994) defined and created reinforcing 
requirements for wall piers, which have been brought into the IBC (ICC, 2003a), but are not in 
ACI 318. Where the wall piers are required for resistance to gravity and lateral loads, 
rehabilitation may be required. Rehabilitation for lateral loads may involve the additional of new 
vertical elements, enhancing of existing elements, or filling in some of the openings. In addition, 
loss of vertical support for the roof adjacent to the panel may be of concern. Rehabilitation for 
vertical loads may be approached by either enhancing the wall element, or providing back-up 
vertical supports. 

Diaphragm Deficiencies 

Due to changes in building code requirements, it is very common for diaphragms in areas of high 
seismic hazard to have inadequate in-plane shear capacity. Regardless, the SEAONC Guidelines 
indicates that diaphragm overstresses have rarely been associated with significant earthquake 
damage. In areas of moderate seismic hazard, this may or may not be a significant deficiency. 
Diaphragm strength and stiffness deficiencies are most often rehabilitated by enhancing the 
existing diaphragm. The addition of new vertical elements to reduce diaphragm span and 
therefore diaphragm shears is also an effective approach, but such an approach is not as 
commonly used. Diaphragm enhancement is addressed in Chapter 22. 

Many California tilt-up buildings have been constructed with solid-sawn roof purlins with sizes 
ranging from 4x12 to 4x16. These framing members may have calculated overstresses under 
existing dead and live loads due to reductions in allowable stresses from in-grade values 
introduced in the 1991 NDS (AF&PA, 1991). When this is the case, removal and replacement of 
roof sheathing is a preferred alternative to overlays in order to keep additional dead load a 
minimum. Diaphragm enhancement using staples may also provide required strength. 
Sometimes, lighter roofing finish materials can be installed where existing roofing is removed, to 
further reduce overstresses. 

Other diaphragm deficiencies include inadequate chord capacity and stress concentrations at 
large diaphragm openings and re-entrant corners. Rehabilitation at re-entrant corners primarily 
involves the provision of adequate chords and collectors. The same is true at large diaphragm 
openings. 

16.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
16.4.1 Enhance Wall-to-Diaphragm Anchorage 
Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses enhanced anchorage of walls into wood sheathed or steel 
deck diaphragms for out-of-plane loads. Cross-ties are a required continuation of the wall 
anchorage system. See Section 22.2.3. 

Wall anchors and cross-ties should be the highest priority for rehabilitation of wood diaphragm 
PC1 buildings. This section illustrates the basic rehabilitation requirements for wall anchorage of 

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wood and steel diaphragm buildings. Refer to the SEAONC Guidelines for more exhaustive 
treatment of detailing. 

The SEAONC Guidelines provides details of code requirements and observed damage over many 
years. Earthquake damage to wall anchorage has been observed not only in older buildings, but 
also in recently constructed buildings, where the type and installation of wall anchors was critical 
to their ability to perform. Just providing anchors is not enough; attention to detailing and field 
installation is critical. 

Description of the Rehabilitation Technique 

Figure 16.4.1-1A illustrates a roof plan for a PC1 building with a wood diaphragm. Figures 
16.4.4-1B through 16.4.4-1D, located as shown on the roof plan, illustrate common rehabilitation 
measures for wall-to-diaphragm anchorage. Also illustrated in Figure 16.4.1-1A are 
subdiaphragms used as part of the diaphragm cross-tie system, described in detail in Chapter 22. 
While subdiaphragms are always used in wood structural panel diaphragms, they are only 
occasionally used in steel deck diaphragms. 

Figures 16.4.1-1B and 16.4.1-1C illustrate anchorage from the west wall to a subdiaphragm 
extending between Lines A and B. Similarly, the east wall is anchored to a subdiaphragm 
between Lines G and H. The primary objective is to create an adequate load path for out-of-plane 
loads acting both into the diaphragm (compression) and away from the diaphragm (tension). This 
type of anchorage is generally provided every four to eight feet on center. The load path into the 
diaphragm includes both wall to framing member anchorage and sheathing nailing to transfer 
loads from the framing member to the sheathing. Figure 16.4.1-1B relies on existing sheathing 
nailing, while Figure 16.4.1-1C adds sheathing nailing. 

Where framing member ends are tight against the wood ledger and the wood ledger is tight 
against the exterior wall, the compression load path can be carried by the framing members. 
Experience has shown that gaps often occur between member ends and the ledger. The 
movement required to close these gaps has been enough to damage the roof sheathing and 
damage devices used for tension anchorage. 

Tie-down devices can provide both the tension and compression load path. In order to do this, 
the device must be rated for both tension and compression loading by the manufacturer, and it 
must be installed in accordance with the manufacturers instructions for compression load. This 
generally involves limiting the unsupported length of the tie-down rod and providing additional 
nuts and washers at the tie-down seat. Very few tie-down devices are currently available that are 
rated for both tension and compression loads. Building codes such as the now require that forces 
and stresses induced by eccentricities in the connection be addressed, and the SEAONC 
Guidelines encourages use of tie-downs placed symmetrically on the purlin or sub-purlin in order 
to minimize eccentric beam loading and encourages use of stiff tie-downs in order to minimize 
deformation demand and possible diaphragm damage resulting from deformation. For girders, 
symmetrical tie-downs are recommended but not as critical as with smaller framing members. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 16  Type PC1: Tilt-up Concrete Shear Walls 

Figure 16.4.1-1A: Roof Plan with Wall Out-of-Plane Anchorage 
for Flexible Wood Diaphragm 

Figures 16.4.1-1A and 16.4.1-1C do not show wall anchors attached to single existing 2x 
subpurlins. It is recommended that anchorage be to a 3x or wider member, or multiple 2xs as 
shown in Figure 16.4.1-1B. While it may be possible for a single 2x4 or 2x6 to be shown 
adequate in calculation, adequate installation and performance are extremely difficult to achieve. 

The wall anchorage system needs to extend across the width of the subdiaphragm. In Figure 
16.4.1-1C, this involves providing extra pairs of tie-down devices between subpurlins across the 
first purlin. Again, both tension and compression load paths are needed. The required 
subdiaphragm depth is determined from the number of nails required to transfer subdiaphragm 
forces into the main diaphragm. Depending on subdiaphragm requirements, additional tie-down 
pairs could be required across more purlins. Subdiaphragm requirements are discussed in 
Chapter 22. 

Details 16.4.1-1B and 16.4.1-1C show work from top of the diaphragm. See Chapter 22 for 
alternates for working from below. Location of access needs to be decided early on in the design 
process and will drive both calculations and detailing of the rehabilitation work. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 16  Type PC1: Tilt-up Concrete Shear Walls 

Figure 16.4.1-1B: Wall Out-of-Plane Anchorage for Flexible 
Wood Diaphragm at Subpurlins  Roofing Not Removed

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 16  Type PC1: Tilt-up Concrete Shear Walls 

Figure 16.4.1-1C: Wall Out-of-Plane Anchorage for Flexible 
Wood Diaphragm at Subpurlins  Roofing Removed 

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Techniques for the Seismic Rehabilitation of 
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Figure 16.4.1-1D illustrates anchorage of the north and south walls to a purlin. In this case, the 
purlin is long enough to extend across the subdiaphragm width (extending between Lines 1-2 and 
3-4), so additional pairs of tie-downs are not needed. As is previous details, both tension and 
compression load paths must be maintained. 

Figure 16.4.1-1D: Wall Out-of-Plane Anchorage 
for Flexible Wood Diaphragm at Purlins 

Figure 16.4.1-2A illustrates a similar roof plan for a PC1 building with a steel deck diaphragm. 
It is important to note in this figure that subdiaphragms (as shown in Figure 16.4.1-1A) are not 
used. Instead, the steel deck provides a continuous cross-tie in the east-west direction, while in 
the north-south direction open web joists provide direct cross-ties across the entire diaphragm 
width at each wall anchor location. This is the primary approach used in new steel deck 
diaphragm construction. Subdiaphragm concepts can be applied to steel deck construction, but 
are not common. 

Figures 16.4.1-2B and 16.4.1-2C provide wall to diaphragm anchorage details. In Figure 16.4.12B, 
wall anchorage forces are transmitted to the steel deck. The deck section, deck edge 
fastening, and deck end splices need to be checked for wall anchorage tension and compression 
forces. Justification of the capacity may be by calculation or testing. The balance of the load path 
also needs to be checked and enhanced as required. In the detail shown, supplemental adhesive 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 16  Type PC1: Tilt-up Concrete Shear Walls 

Figure 16.4.1-2A: Roof Plan with Wall Out-of-Plane Anchorage 
for Flexible Steel Diaphragm 

anchors to the wall and a second ledger angle are provided. Prying action in the steel ledger 
angle needs to be considered in determining wall anchor forces. 

Figure 16.4.1-2C shows wall anchorage to a steel open web joist. The wall anchorage connection 
may be through the joist seat in new construction; however, a supplemental anchor to the joist is 
likely needed in rehabilitation. The joist must be checked for wall anchorage forces and any 
applicable eccentricities. Details in Chapter 22 address joist to joist connections to complete the 
cross-tie. 

Compression forces can be carried in tie-down devices, if rated for compression by the 
manufacturer. Unsupported tie-down rod lengths must be kept short, and additional nuts and 
washers are needed to transfer compression. Again, wall anchorage loads need to be transferred 
into the decking. In this case existing welding or screwing of the decking is relied upon. If this 
fastening is not adequate, the roofing will need to be removed to allow additional fastening. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 16  Type PC1: Tilt-up Concrete Shear Walls 

Figure 16.4.1-2B: Wall Out-of-Plane Anchorage for Flexible Steel Diaphragm 

 to Decking for Load Parallel to Flutes 
Design and Detailing Considerations 

Research basis: No research applicable to the performance or adequacy of enhanced anchorage 
methods has been identified; however, the demands created in flexible diaphragms have been 
studied by Fonseca, Wood and Hawkins (1996); Hamburger and McCormick (1994); Ghosh and 
Dowty (2000); and Freeman, Searer, and Gilmartin (2002). 

As discussed in Section 16.1, even wall anchorages constructed or rehabilitated in the 1980s and 
early 1990s were observed to have been damaged in the 1994 Northridge earthquake. The reader 
is referred to the extensive discussion in the SEAONC Guidelines for design and detailing 
considerations and lessons learned. 

Anchor type and installation: A variety of proprietary anchors are available for anchorage to 
existing concrete walls. Both manufacturer literature and ICC Evaluation Service reports should 
be consulted for information on conditions of use, allowable loads, and installation and 
inspection requirements. It is important to make sure that the anchor type is appropriate for the 
material to which it will be connected and is approved for seismic loads. The diameter of drilled 
holes is specified in installation requirements for each anchor type; variation from this size often 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 16  Type PC1: Tilt-up Concrete Shear Walls 

Figure 16.4.1-2C: Steel Open Web Joist Anchorage to Exterior Wall 

leads to inadequate anchor capacity. Most manufacturers have caulking gun-like devices that 
make field placement of epoxy fairly simple and automatically mix two-part adhesives. 
Generally, these types of adhesives provide more than adequate strength, and there is no need to 
use more complicated high-strength adhesive types. The cleaning of holes prior to placing 
adhesive anchors is paramount for anchor capacity. When not well cleaned, the anchors can pull 
out at a small fraction of the design load. It is common to pull-test a portion of the adhesive 
anchors to verify adequate installation. The pull test load is usually in the range of one to two 
times the tabulated allowable stress design tension load. The bridge used for testing generally 
makes a concrete cone pull-out failure unlikely. The test load should not be near yield load for 
bolts or adhesive pull-out failure loads. 

Anchors added in rehabilitation will often have to work in combination with existing cast-in 
anchors. In order to allow load sharing, anchorage of similar stiffness is desirable. This is often 
best achieved with adhesive anchors. In addition, anchorage provisions for new buildings have 
moved towards having the attachment to concrete be capable of developing the yield capacity of 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 16  Type PC1: Tilt-up Concrete Shear Walls 

the steel anchor in order to promote ductile connection behavior. Again, this is best achieved 
with adhesive anchors. Anchor types other than adhesives should be carefully evaluated. 

Cost, Disruption and Construction Considerations 

When rehabilitation work is undertaken on the roof diaphragm of a PC1 building, it is important 
that the cost and the preferred location for work (from the underside or top of roof) take into 
account the combination of work, rather than considering one piece at a time. If several 
diaphragm measures will be undertaken, it will quickly become cost-effective to remove the roof 
and allow work from the top. This is particularly true if a steel deck requires several 
rehabilitation measures. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique, other than the use of 
proprietary connectors and adhesive anchors as part of the assemblage. 

16.4.2 Enhance Beam and Girder Connection to Supporting Elements 
Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses enhancement of girder gravity load connections to tilt-up 
walls. While primarily intended to carry gravity loads, these connectors should also be adequate 
to resist wall out-of-plane loads. 

Description of the Rehabilitation Technique 

Tilt-up walls may have pilasters supporting girder loads. In older tilt-up buildings, these are often 
cast-in-place pilasters, while in newer buildings pilasters are cast as part of the tilt-up panel. The 
pilaster acts as a wall stiffener, allowing the wall to span both horizontally and vertically under 
out-of-plane earthquake loading. This attracts more load to the pilaster, and the top of pilaster 
reaction for out-of-plane loading will be higher than at a typical wall anchor. The gravity load 
connection generally also serves as the girder-to-pilaster connection for out-of-plane loads. Two 
deficiencies are common with this connection: 1) inadequate confinement around anchor bolts 
embedded in the pilaster top and 2) inadequate connection to the girder. Tension loads on the 
connector have led to splitting of the column top, pulling away the wedge of concrete in front of 
the anchor bolts. In recent codes, the placement of three closely spaced ties at anchor bolts has 
been required to reinforce across the anticipated concrete crack. Where added ties have not been 
provided, a collar around the pilaster top can provide external reinforcement (Figure 16.4.2-1). 
The collar should be relatively stiff to minimize splitting of the concrete before load is taken up. 
The second issue is inadequate connection to the wood girder, including bolt capacity and 
placement. Where the girder seat connection to concrete is adequate or can be enhanced, 
inadequate bolt capacity can be mitigated with addition of bolt tabs (Figure 16.4.2-2). 
Alternately, it is possible to use new wall anchors from the girder to the wall, bypassing the 
girder seat (Figure 16.4.2-3). Again, this connection must be stiff. 

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Techniques for the Seismic Rehabilitation of 
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Figure 16.4.2-1: Enhanced Girder Connection  Collar at Pilaster 

Adapted From SEAONC (2001) 

Girders that are supported directly on a flat wall panel using a steel U-bracket bolted or welded 
to the panel (Figure 16.4.2-3) will also attract wall out-of-plane forces. As is true with wall 
pilasters, a girder and U-bracket are likely to provide a stiffer load path for wall out-of-plane 
loads than adjacent anchors. For this reason, use of a wall anchorage force greater than used for 
adjacent anchors is encouraged. The girder connection should have the ability to resist wall 
anchorage loads in combination with gravity loads. Anchorage of the bracket to the panel will 
often be adequate for both gravity and lateral loads; however, the bracket attachment to a wood 
girder will often not have the quantity or placement of bolts required for tension loads. Addition 
of steel tabs and bolts will add capacity and place bolts where end distances are adequate for 
tension loads. Where the steel connection to the concrete is not adequate, the out-of-plane 
anchor might bypass the existing connection and connect the girder directly to the wall. Figure 
16.4.2-2 shows two approaches, one with a tie-down on each side of the girder and a second with 
a tie-down on the girder bottom. The out-of-plane wall anchor should be as stiff as possible to 
minimize damage to the gravity connection. 

Design and Detailing Considerations 

Research basis: No research applicable to this rehabilitation measure has been identified. 

See Section 16.4.1 wall anchorage and the SEAONC Guidelines for additional detailed 
discussion. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 16  Type PC1: Tilt-up Concrete Shear Walls 

Figure 16.4.2-2: Enhanced Girder Connection at U-hanger 

Adapted From SEAONC (2001) 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique, other than the use of 
proprietary connectors as part of the assemblage. 

16.4.3 Enhance Anchorage at Base of Tilt-Up Panels 
Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses the addition or enhancement of connections between the 
tilt-up wall panel and foundation or adjacent slab-on-grade to resist in-plane shear and 
overturning forces and out-of-plane wall anchorage forces. 

Description of the Rehabilitation Technique 

Rehabilitation of base connections in PC1 buildings for in-plane and out-of-plane shear loads is 
most commonly accomplished by addition of steel angles and adhesive anchors between the wall 
panel and adjacent slab-on-grade. This is illustrated in Figure 16.4.3-1. In some instances, the 
slab-on-grade may not have been thickened adjacent to the tilt-up panel. When this is the case, it 
may be necessary to remove and recast a thicker pour strip in order to get adequate anchorage. 
The connection shown would flex if the wall were to uplift. Where uplift connection capacity is 
required, a direct tension connection of the wall to the footing below is recommended. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 16  Type PC1: Tilt-up Concrete Shear Walls 

Figure 16.4.2-3: Enhanced Girder Connection at Pilaster 

Adapted From SEAONC (2001) 

Variations in base conditions include 1) older PC1 buildings that may not have any doweling 
because friction was relied on to resist forces at the base of the panel and 2) welded connections 
between cast-in embeds in the wall panel and slab, similar to PC2 wall panel connections. 

Design and Detailing Considerations 

Research basis: No research applicable to this rehabilitation technique has been identified. 

The SEAONC Guidelines provide discussion of a variety of possible existing conditions, changes 
in code requirements, and implications for retrofit. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique, other than the use of 
proprietary adhesive anchors as part of the assemblage. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 16  Type PC1: Tilt-up Concrete Shear Walls 

Figure 16.4.3-1: Enhancement of Tilt-up Panel Base Connection 

16.5 References 
Extensive reference and bibliography listings are included in SEAONC (2001). The following 
only includes references specifically cited in this chapter. 

ACI, 2005, Building Code Requirements for Structural Concrete, American Concrete Institute, 
Farmington Hills, MI. 

ACI-SEASC Task Group on Slender Walls, 1982, Report of the Task Group on Slender Walls 
(Green Book), American Concrete Institute, Southern California Chapter and the Structural 
Engineers Association of Southern California, Los Angeles, CA. 

AF&PA, 1991, ANSI/NFoPA NDS-1991 National Design Specification for Wood Construction, 
American Forest and Paper Association, Washington, D.C. 

City of Los Angeles, 2002, Chapter 91, City of Los Angeles Building Code, Los Angeles, CA. 

EERI, 1996, Supplement C to Volume 11, Northridge Earthquake of January 17, 1994, 
Reconnaissance Report, Earthquake Spectra, Volume 2, EERI, Oakland, CA. 

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Techniques for the Seismic Rehabilitation of 
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EERI, April 1995, Guam Earthquake of August 8, 1993 Reconnaissance Report, Earthquake 
Spectra, Supplement B to Volume 11, Earthquake Engineering Research Institute, Oakland, Ca. 

FEMA, 2004, NEHRP Recommended Provisions for Seismic Provisions for New Buildings and 
Other Structures, FEMA 450, Federal Emergency Management Agency, Washington, D.C. 

Freeman, S., S. Searer and U. Gilmartin, 2002, Proposed Seismic Design Provisions For Rigid 
Shear Wall / Flexible Diaphragm Structures, Proceedings of the Seventh U.S. National 
Conference on Earthquake Engineering, EERI, Oakland, CA. 

Fonseca, F., S. Wood and N. Hawkins, 1996, Measured Response of Roof Diaphragms and 
Wall Panels in Tilt-Up Systems Subject to Cyclic Loading, Earthquake Spectra, Volume 12, 
Number 4, Earthquake engineering Research Institute, Oakland, CA. 

Ghosh, S.K. and S. Dowty, 2000, Anchorage of Concrete or Masonry Walls to Diaphragms 
Providing Lateral Support, Draft 2000, Not Published. 

Hamburger, R.O. and D. McCormick, 1994, Implications of the January 17, 1994 Northridge 
Earthquake on Tilt-up Wall and Masonry Wall Buildings with Wood Roofs, Proceedings of the 
1994 Convention of the Structural Engineers Association of California, Sacramento, CA. 

ICBO, 1994, Uniform Building Code, International Conference of Building Officials, Whittier, 
California. 

ICBO, 2001, Guidelines for Seismic Retrofit of Existing Buildings (GSREB), International 
Conference of Building Officials, Whittier, CA. 

ICC, 2003a, International Building Code, International Code Council, Country Club Hills, IL. 

ICC, 2003b, International Existing Building Code, International Code Council, Country Club 
Hills, IL. 

SEAONC (Structural Engineers Association of Northern California), 2001, Guidelines for 
Seismic Evaluation and Rehabilitation of Tilt-Up Buildings and Other Rigid Wall/Flexible 
Diaphragm Structures, SEAONC, San Francisco, California. 

SEAOSC (Structural Engineers Association of Southern California), 1979, Recommended Tilt-up 
Wall Design (Yellow Book), SEAOSC, Los Angeles, CA. 

Tilt-up Concrete Association, 2004, Tilt-up Construction and Engineering Manual, Sixth 
Edition, Tilt-up Concrete Association, Mount Vernon, IA. 

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Chapter 17 - Building Type PC2: Precast Concrete Frames 
with Shear Walls 

17.1 Description of the Model Building Type 
Buildings designated as PC2 include wide ranging combinations of precast and cast-in-place 
concrete elements. Precast members may be limited to a floor system of hollow core or T-beam 
construction, or may include all elements of the gravity and lateral load systems. For this chapter, 
Building Type PC2 includes concrete wall or frame buildings in which any of the horizontal or 
vertical elements of the lateral load system are of precast concrete, except for flexible diaphragm 
buildings which are addressed as Building Type PC1 in Chapter 16. 

Figure 17.1-1: Building Type PC2: Precast Concrete Gravity Frames with Shear Walls 

Extensive use of hollow core floor systems in buildings with concrete and masonry walls in 
southern regions of the United States makes this the single largest group of PC2 buildings. 
Parking garages (used exclusively for parking rather than mixed use) represent the next largest 
group of PC2 buildings and a substantial portion of the current PC2 building inventory in the 

U.S. The PC2 building type has also been used for a variety of other occupancy types in the U.S. 
and internationally, including mid-rise office, hotel, and residential buildings, low-rise 
residential, commercial, and prison buildings. 
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Over the past decade, significant effort has been devoted to development and testing of precast 
ductile moment frame systems through the PRESSS (Priestley et al., 1999). These systems are 
not addressed in this chapter due to their very recent development and state-of-the-art detailing. 

Gravity-Carrying Load Systems 

Special attention is needed to PC2 buildings in which concrete frames (beams, girders, and 
columns or moment frames) resist gravity load, or a combination of gravity and seismic load. 
Very important to the performance of all concrete buildings with frames, including PC2 
buildings, is the lack of ductile detailing in concrete columns not designated as part of the 
seismic force-resisting system. These columns in many instances do not have confining steel 
adequate to accommodate the drift imposed by the seismic force-resisting system and as a result 
fail through longitudinal bar buckling and concrete crushing. Requirements for estimation of 
building drift have changed over time, and understanding of potential building deflection has 
improved with each observation of earthquake performance. As a result, it is important to revisit 
the ability of non-ductile columns to accommodate estimated drifts, even if they were checked 
when initially designed. In some precast buildings, the division of initial design responsibility 
between one engineering firm for the gravity load system and a second firm for the seismic 
force-resisting system may have contributed to inability to accommodate estimated building 
deflections. In earthquake performance to date, diaphragm deflections have been a large 
contributor to deflection of non-ductile gravity systems. Vertical elements, and most particularly 
moment frames, could also contribute significantly to gravity system deflection. 

Following the 1994 Northridge earthquake, column detailing requirements and methods of 
estimating building deflection for purposes of gravity frame design were modified in codes and 
standards. Concrete buildings constructed recently in areas of high seismic hazard should 
perform significantly better than those designed under older provisions. Research continues to 
develop a better understanding of sources of diaphragm deflection. 

When considering the ability of gravity load-carrying systems to accommodate building 
deflection, a related issue of importance is proper accounting for column stiffness and restraint in 
analysis. This, again, is a concern for all concrete buildings including PC2 buildings. Short 
columns will attract higher forces due to increased stiffness and have been seen to fail as a result. 
Short columns can be created accidentally due to inadequate separation of the column from 
nonstructural components such as guardrails. In addition, systematic problems with short 
columns can occur at parking garage ramps. Analysis models need to pay special attention to 
these and other sources of shortened columns or columns with increased end fixity. 

PC2 buildings with gravity and lateral loads supported exclusively by structural walls do not 
have the same issues of deflection of non-ductile columns. Connections tend to be the primary 
issue of importance to these systems for both gravity and seismic load systems. 

Shear Walls and Frames 

Building Type PC2 may have a lateral force-resisting system of concrete shear walls or moment 
frames, cast-in-place or precast. In PC2 buildings, critical behavior of shear walls is generally 
governed by connections including: diaphragm to shear wall, shear wall above to shear wall or 
foundation below, and interconnection of shear walls within a story. In PC2 buildings with 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

precast frames, field connections within the frame are critical to performance, as is ductile 
detailing. Connection practice has varied widely over time and by geographic region. 

Floor and Roof Diaphragms 

In California, precast floor T-beams or hollow core planks are covered by a cast-in-place topping 
slab, reinforced to provide diaphragm action. These toppings need to be clearly differentiated 
from topping slabs plant-applied to individual precast members, which do not serve the same 
function of structurally interconnecting adjacent members. Welded connections between 
embedded inserts or plates may also be used to aid in alignment of members during erection, but 
are generally not relied on for diaphragm action. Reinforcing bars are often added in the cast-inplace 
topping slab to act as diaphragm chords and collectors, and welded wire fabric is used to 
provide shear reinforcement. 

In other areas of the United States, common methods of joining floor sections include use of 
grouted hollow core joints (grout placed in the joint between two adjacent panels, relying on 
adhesion and/or friction for shear transfer) or welded insert plates. Cast-in-place topping slabs 
are not commonly used. In some areas outside the U.S., hollow core planks are installed with no 
connection or grouting between adjacent planks. 

As per the discussion of gravity load systems, deflection of the diaphragm system has been seen 
as a significant contributor to building deflection is past earthquakes. Discussion of diaphragm 
behavior and rehabilitation can be found in this chapter, Chapter 20 for masonry wall buildings, 
and Chapter 22 for detailed discussion of diaphragm rehabilitation. 

Parking Structure Issues 

Parking structure PC2 buildings have unique characteristics that deserve specific discussion, 
some applicable to parking structures regardless of structural system and others specific to 
precast construction. These issues include the following. 

Many parking structures have large plan areas, and considerations of security and 
restraint against temperature, creep, and shrinkage movement lead to concentration of the 
shear walls at the building perimeter near the center of each side (Figure 17.2-1). This 
configuration leads to long diaphragm spans with significant shear, moment, and 
collector demands. With these high demands, it is possible for the diaphragm, rather than 
the vertical elements to control building dynamic behavior. This is of concern in all 
systems, but particular in precast systems due to the lower level of inherent diaphragm 
continuity. 
Compared to other building uses, parking garages have greatly minimized finish and 
cladding systems, resulting in low levels of nonstructural damping and energy 
dissipation. 
Ramps in parking structures may act as tension and compression struts between floors, 
resulting in demands not anticipated during design. This behavior can be avoided by 
inclusion of seismic joints at one end of each ramp; however, seismic joint detailing is 
difficult to accommodate in precast concrete construction, making use of a fixed 
connection more likely. Unless the effect of the ramp is specifically considered in 
analysis, force transfer through the ramp can result in seismic forces bypassing the 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

intended resisting system and significant redistribution of forces in the diaphragms and 
vertical elements. An analytical study of ramp effects discussed in Lyons, Bligh, 
Purlinton, and Beaudoin (2003) suggests that while the effect of ramps is significant in 
moment frame buildings, it is less significant and can often be managed in design of 
shear wall buildings. The effect of ramps should be considered in evaluation and 
rehabilitation of parking structure buildings. 

Figure 17.2-1: Plan of Common Parking Structure Configuration 

17.2 Seismic Response Characteristics 
PC2 buildings occur with a wide range of vertical element types. In most cases, the vertical 
element type will dictate the building seismic response: shear wall buildings will have short 
period response, while frame buildings will have a longer period. In PC2 buildings, stiff 
diaphragm behavior will generally be intended. Parking structure PC2 buildings with long 
diaphragm spans, however, have been observed to have inelastic behavior concentrated in the 
diaphragms rather than vertical shear wall or frame elements. To date, this has been brittle 
behavior resulting in premature diaphragm failure; however, with development of proper 
detailing it may be possible to achieve stable long-period diaphragm behavior. 

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17.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
Construction of PC2 buildings in areas of high seismic hazard in the U.S. has been of limited 
quantity and recent compared to most other building types, resulting in limited opportunities to 
observe earthquake performance; the City of Los Angeles and SEAOSC (1994) and EERI (1996) 
reported on performance of one group of PC2 buildings following the 1994 Northridge 
earthquake. Out of an estimated 100 parking garages (precast and cast-in-place) in heavily 
shaken areas in the Northridge earthquake, eight had partial collapses, and an additional 20 had 
at least 25% damage (City of Los Angeles and SEAOSC, 1994). The task force looked at 
approximately 30 structures; of 26 structures with damage, approximately half contained some 
precast elements (Mooradian, 2005). 

Within limited experience to date, life-safety performance of other PC2 buildings in the U.S. has 
been good; however, performance in other countries has identified concerns that could be 
applicable to U.S. construction. See below for general discussion and Table 17.3-1 for a detailed 
compilation of common seismic deficiencies and rehabilitation techniques for the PC2 building 
type. 

Global Strength and Stiffness 

Insufficient in-plane shear wall strength and stiffness are possible seismic deficiencies in PC2 
buildings and particularly in parking garages where shear wall length is generally limited. 
Rehabilitation to address inadequate shear wall strength and stiffness can include addition of new 
vertical elements or strengthening of existing elements, as summarized in Table 17.3-1. Addition 
of and enhancement to elements in PC2 buildings is very similar to other concrete building 
types; however, several additional cautions are in order. First, the configuration of existing 
precast members, including cast-in voids and prestressing tendons must be carefully studied to 
allow connection of new elements to existing construction. Second, precast and post-tensioned 
systems are configured to minimize damaging effects of movement due to temperature variation, 
shrinkage and creep; these effects should be considered in the addition of new vertical elements. 

Insufficient in-plane moment frame strength is a possible seismic deficiency in PC2 buildings 
and particularly of concern where the frame might not have been initially designed for seismic 
loads. Where strength is a concern, it is likely that stiffness, connections, and ductile detailing 
will also be inadequate and that major addition or enhancement of vertical elements is required. 

Configuration 

Torsional irregularities can lead to possible seismic deficiencies in PC2 buildings, as in any 
other building type, increasing deformation demand in local portions of the structure. One of the 
parking garages investigated by the City of Los Angeles and SEAOSC Task Force (City of Los 
Angeles and SEAOSC, 1994) had shear walls on three sides and an open front of the fourth side. 
The report speculates that excessive deflection at the open-front allowed girders to move 
sideways off of supporting columns, resulting in total collapse. The torsionally irregular building 
configuration appears to have contributed to collapse, along with inadequate diaphragm stiffness 
and component connections. Torsional deficiencies are most directly addressed through the 
addition of new vertical elements, as indicated in Table 17.3-1. Design to accommodate the 
concentrations of force and deformation demand may be an alternative. 

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Techniques for the Seismic Rehabilitation of 
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Table 17.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for PC2 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce 
Demand 
Remove Selected 
Components 
Global 
Strength 
Insufficient in-plane 
strength of shear 
walls or frames 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [12.4.2] 
Fiber composite wall 
overlay [13.4.1] 
Concrete wall 
overlay [21.4.5] 
Infill openings 
Global 
Stiffness 
Inadequate stiffness 
of shear walls or 
frames 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [12.4.2] 
Fiber composite wall 
overlay [13.4.1] 
Concrete wall 
overlay [21.4.5] 
Infill wall openings 
Configuration Torsional 
irregularity 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [12.4.2] 
Incompatible 
deformation of 
building sections 
Provide seismic 
separation of 
portions with 
different 
behavior. See 
general discussion 
of seismic 
separation. 
Distance between 
shear walls too large 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [12.4.2] 
Load Path Inadequate force 
transfer, diaphragm 
to shear wall, shear 
wall above to shear 
wall below, shear 
wall to foundation 
Enhance anchorage 
[17.4.2] 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

Table 17.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for PC2 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements 
Reduce 
Demand 
Remove Selected 
Components 
Load Path 
(continued) 
Inadequate 
connection of beam 
or girders to 
supporting elements 
Supplemental 
vertical supports 
[21.4.11] 
Enhance anchorage 
Inadequate 
collectors 
Add collector 
[17.4.1] 
Enhance existing 
collector [17.4.1] 
Component 
Detailing 
Gravity columns 
inadequate to 
accommodate drift 
Reduce building 
drift to level 
acceptable for 
gravity elements  
see global stiffness 
Enhance column 
ductility with 
jacketing [12.4.4] 
Remove or 
reconfigure 
portions of 
structure creating 
short columns 
[17.4.3] 
Inadequate wall 
strength 
See Global Strength See Global Strength 
Inadequate frame 
connection detailing 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [12.4.2] 
Enhance existing 
frame connections 
Diaphragms Inadequate strength 
and/or stiffness 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [12.4.2] 
Enhance shear 
transfer within 
diaphragm [22.2.11] 
Inadequate shear 
transfer to walls 
See load path 
Inadequate chord 
capacity 
Add chord [17.4.1] Enhance existing 
chord [17.4.1] 
Excessive stresses at 
openings and 
irregularities 
Enhance diaphragm 
detailing 
Re-entrant corners Enhance diaphragm 
detailing 
Foundations See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation measures. 

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Techniques for the Seismic Rehabilitation of 
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One of the observations from the Northridge earthquake was that performance of parking 
garages decreased as the distance between supporting shear walls increased. Long diaphragm 
spans resulted in high shear and flexure demands on the diaphragms, causing yielding and 
fracture of diaphragm reinforcing (City of Los Angeles and SEAOSC, 1994; and Wood, Stanton 
& Hawkins, 2000). This can be considered either a configuration deficiency or a diaphragm 
deficiency. The addition of vertical elements will contribute significantly to reduction of demand 
and deflection. See global strength and stiffness discussion. Diaphragm shear capacity, chords, 
and collectors can also be enhanced, as discussed in the diaphragm deficiency section. 

Where PC2 buildings are comprised of several sections separated by seismic joints, movement 
can be incompatible and separation inadequate. 

Load Path 

Deficiencies in the load path connections between diaphragms and vertical elements, story to 
story and wall to foundation are of significant concern in every type of PC2 building. 
Connection and load path detailing in the existing PC2 building stock is thought to range from 
systems with no positive connections at all, to potentially brittle welded connections, to recently 
developed connections and systems that may allow better performance than cast-in-place 
buildings. Good performance was reported for limited examples of low-rise PC2 shear wall 
buildings in the Northridge earthquake (Iverson and Hawkins, 1994) and the Kobe earthquake 
(Ghosh, 1995). In the Guam earthquake (EERI, 1995) damage to the shear wall to foundation 
connections in a mid-rise hotel building caused extensive local spalling, apparently related to 
eccentricities within the connections. In the Armenia earthquake (EERI, 1989) low to mid-rise 
large panel precast buildings performed well. The good performance was attributed to floor 
panels that spanned to bearing walls on all four sides, and to the redundancy of these systems. In 
contrast, the connections between precast column members in frame buildings were very 
vulnerable, likely due to eccentricities introduced in site modifications to longitudinal bar splices 
and poor column confinement. Overall load path connections are important to the performance of 
PC2 buildings, and attention to detail and eccentricities is important. Rehabilitation of load path 
connections in shear wall type buildings will generally involve external mechanical connections. 

While not as obvious an issue in PC2 buildings as in PC1, as part of load path considerations 
concrete wall panels must be adequately connected to resist out-of-plane forces. The connections 
that are transferring diaphragm forces to the shear walls are generally also used to resist out-ofplane 
forces, so rehabilitation of these connections should consider both demands. 

The connection between girders and supporting columns and other similar connections may 
require rehabilitation in order to provide a continuous load path. The movement of a girder off of 
the supporting column due to building deflection is the most obvious concern. It has additionally 
been postulated that, in the Northridge earthquake, some gravity members may have been pulled 
free of supports due to high vertical accelerations or vibrations. At one time, it was common to 
have fairly heavy connections between girders and columns at the point of bearing. Connection 
design would likely have controlled by calculated forces, and welded connection behavior could 
be brittle. It is now more common to minimize or eliminate connection between the column and 
girder at the bearing point and resulting restraint of support movement and to rely instead on the 
column and girder each being doweled into the diaphragm system. This approach minimizes 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

unintended restraint and resulting damage. As in new construction, rehabilitation of this 
connection would best be accomplished by connection of each member into the diaphragm 
system. This approach is best suited to the intended behavior of the building system. Alternate 
approaches could include use of restraint cables, as is common in bridge rehabilitation or use of 
secondary vertical supports, as is common in rehabilitation of unreinforced masonry buildings 
(Section 21.4.10). There is no broad consensus on the contribution of vertical accelerations or 
vibrations to damage of PC2 buildings in the Northridge earthquake. In locations where vertical 
thrust is a concern, rehabilitation measures could specifically take vertical demands into 
consideration. 

In some precast systems, design of connections between elements has been based on the concept 
that the precast system emulates performance of an equivalent cast-in-place system. In frame 
members, the splicing of reinforcing steel is key to this performance. Over the years, the 
understanding of demands on splices and adequacy of splice technologies under earthquake 
loading has changed, as well as understanding of desirable locations of splices and controlling 
behavior modes within concrete frame systems, both precast and cast-in-place. Rehabilitation of 
connections between members in precast frames is difficult, and addition of vertical elements to 
limit frame drift may be a more practical solution. 

Component Detailing 

The inability of columns in gravity load systems to accommodate building drift has been pointed 
out as a significant deficiency in earlier discussion. Rehabilitation approaches include either 
adding additional vertical elements to reduce drift or enhancing columns with fiber-reinforced 
polymer wraps or similar systems to allow ductile behavior. These rehabilitation measures are 
discussed further in Chapter 12. The City of Los Angeles/SEAOSC report (1994) identified 
column wrapping as the only rehabilitation measure for parking garages that could be 
recommended as both practical and economical. Rehabilitation for the related issue of short 
columns can sometimes be as simple as creating an adequate joint between the column and the 
incidental restraint. Where this is not possible, reducing drift or enhancing column ductility are 
recommended rehabilitation approaches. 

Diaphragm Deficiencies 

Insufficient in-plane strength and stiffness of diaphragms are of significant concern over a range 
of precast systems. 

In parking garages with long diaphragm spans, insufficient shear strength has been identified as a 
likely contributor to poor performance of PC2 parking garages in the 1994 Northridge 
earthquake (Wood, Stanton & Hawkins, 2000), where pre-earthquake cracking of cast-in-place 
topping slabs occurred along the joists between T-beams. Analytical studies identified two 
strength-related issues that had not been considered previously. First, the pre-earthquake 
cracking of the topping slab along T-beam joints meant that the topping slab concrete was not 
contributing shear strength, leaving the reinforcing behavior acting in a shear-friction mode. 
Second, the limited strain capacity of the welded wire fabric reinforcing was being exceeded in 
commonly observed pre-earthquake crack widths, leaving the reinforcing vulnerable to brittle 
fracture during earthquake loading. Further, inadequate performance of chord and collector 
reinforcing in topping slabs has also been identified as a deficiency contributing to damage. In 

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Techniques for the Seismic Rehabilitation of 
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the Northridge earthquake, reinforcing bars serving as chords and collectors in cast-in-place 
topping slabs appear to have yielded and subsequently buckled (City of Los Angeles/SEAOSC, 
1994). This behavior is a combination of strength and stiffness issues. In order to limit 
diaphragm deflection, maintain the integrity of the chord and collector members, and have the 
vertical elements control building dynamic behavior, it would be desirable to not have the chord 
and collector reinforcing yield. Since the Northridge earthquake, ACI 318 (ACI, 2005) and the 
building codes have taken initial steps towards addressing observed problems. Nakaki (1998) 
observed that the prescriptive steps taken by ACI 318 did not always result in improved 
behavior. Nakaki proposed simplified approaches to estimation of force and deformation demand 
for the purpose of ensuring elastic diaphragm design. This work has been incorporated into an 
appendix chapter of the NEHRP Provisions (FEMA, 2004) for untopped diaphragms. 

In diaphragms without cast-in-place topping slabs, connections between adjacent planks or T-
beams often use embedded steel plates with field-welded connections, or grout connections. 
Questions arise as to the pre-earthquake adequacy of these connections. Welded connections are 
often used to correct differences in camber between adjacent members during initial erection and 
are often stressed by moving vehicle point loads and shrinkage and creep movement of the 
building. Observations of connections suggest that reduced capacity prior to earthquake loading 
may be common. This combines with a changing understanding of earthquake demands on the 
connection and the interaction of shear demands and deformation due to flexural or tension 
loading. The complete lack of connection between hollow core floor planks within diaphragms 
appears to have been a primary contributor to collapse of nine-story residential precast concrete 
frame buildings in the 1988 Armenia earthquake (EERI, 1989). 

A significant integrated analytical and experimental research program is currently underway to 
develop a comprehensive design methodology for precast concrete diaphragm systems. The 
project intends to address the discrepancy between current design practice, based on inelastic 
behavior concentrating in vertical elements and observed performance in which substantial 
inelastic behavior has occurred in diaphragms (Wan et al., 2004; and Naito and Cao, 2004). The 
project proposes to determine force and deformation demands required for design, connection 
details to support the performance, and address deformation relative to the gravity load-carrying 
system. This information will be invaluable for both new design and rehabilitation. Testing will 
include individual connections, joints, and half-size components. Analytical modeling of full 
buildings is being used to identify critical demands. Of particular interest is the simultaneous 
occurrence of shear and tension or compression on connections normally considered to carry 
only shear. Published information to date (Naito and Cao, 2004) provides a database of 
connector properties from existing literature and suggests a simplified analysis model based on 
initial finite element testing. Additional information should be available over the next several 
years. 

Rehabilitation of diaphragm chord and collector members is reasonably practical due to the 
focused locations of work. Where possible, it is easiest to add reinforcing steel collectors on the 
top of the floor system in new cast concrete curbs. Where chords and collectors need to occur at 
building interior locations where traffic must cross, more complex solutions are required. 
Unstressed post-tensioning tendons may be a desirable alternative to rebar in some locations; 
however, it must be kept in mind that stresses must be kept low in order to minimize 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

deformation, so the high strength does not give particular advantage. In undertaking 
rehabilitation of the chord and collector members, it must be acknowledged that there is lack of 
consensus on the interaction of shear and flexure demand in these buildings. Care must be taken 
not to induce brittle shear failure of the floor diaphragm as a result of flexural strengthening. 

Rehabilitation of inadequate shear capacity is significantly more difficult. Most precast floor 
systems will have little capacity to support vertical load from additional topping slab thickness, 
and removal of existing toppings over large areas is not practical. Research conducted on 
connections between precast wall panels can be applied to connections between precast 
diaphragm members. Research by Pantelides, Volnyy, Gergeley, and Reaveley (2003), discussed 
in relation to load path connection in Section 17.4.2, could be applied to precast diaphragms. In 
addition a research program is currently being conducted by the Precast/Prestressed Concrete 
Institute investigating development of ductile panel-to-panel connections in precast diaphragms. 
See Section 22.2.11 for further discussion. 

Rehabilitation for large openings and re-entrant corners in PC2 buildings involves providing 
adequate chord and collector members in the vicinity of the opening or corner. Rehabilitation 
methods discussed in Section 17.4.1 are applicable. Framing bays at ramps in parking garages 
may need to be treated as openings for purposes of diaphragm design. 

17.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
To date, very little rehabilitation of PC2 buildings has occurred in the U.S. As a result, the 
following discussion of rehabilitation measures draws from limited available research, suggested 
details for new PC2 construction, and application of rehabilitation techniques for concrete 
buildings to the specific configurations of precast elements. 

17.4.1 Add or Enhance Collector or Chord in Existing Precast Diaphragm 
Deficiencies Addressed by Rehabilitation Technique 

This rehabilitation technique addresses deficient diaphragm boundary members  chords and 
collectors at diaphragm boundaries and at interior openings and re-entrant corners. 

Description of the Rehabilitation Technique 

In diaphragms with cast-in-place topping slabs, existing chords and collectors are likely to be 
reinforcing bars at the edge of the cast-in-place slab and in line with shear walls. The most 
practical rehabilitation technique is addition of structural steel sections or reinforcing bars at the 
diaphragm boundary locations. Where boundary members occur at the perimeter of parking 
structures, it may be possible to encase the steel sections or reinforcing in new concrete curbs on 
top of the existing deck, as shown conceptually in Figure 17.4.1-1. Where boundary members 
extend across the floor plate where foot or vehicle traffic will occur, alternate chord and collector 
locations are required, as shown conceptually in Figures 17.4.1-2 and 17.4.1-3. For both chords 
and collectors, shear transfer capability between the boundary member and the structural 
diaphragm needs to be provided. Adhesive anchors or reinforcing dowels are the most likely 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

Figure 17.4.1-1: Added or Enhanced Chord or Collector at Floor Perimeter 
with Cast-In-Place Topping Slab 

Figure 17.4.1-2: Added or Enhanced Chord or Collector at Floor Interior 
with Cast-In-Place Topping Slab 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

Figure 17.4.1-3: Added or Enhanced Chord or Collector at Floor Interior with 
Cast-In-Place Topping Slab 

methods of attachment. Chord and collector splices must also be detailed. For collectors, transfer 
of the collector force to the vertical elements of the lateral force-resisting systems is required. 
Figure 17.4.1-4 shows a conceptual approach where the chord/collector curb runs by the face of 
the shear wall and dowels in over the full shear wall length. Figure 17.4.1-5 illustrates 
considerations when collector steel is to be doweled into the end of an existing shear wall, which 
may not be advisable. 

In diaphragms without cast-in-place topping slabs, reinforcing in wall panels, floor panels, or 
beams will likely serve as chord and collector reinforcing. Where these floor panels or beams are 
precast, the connection between members is likely to be the weak link in chord and collector 
capacity. Figures 17.4.1-6 and 17.4.1-7 show the concept of added steel angles used as chords 
and collectors. The steel angle also serves as the connection between the wall and diaphragm for 
in-plane and out-of-plane forces. 

Design Considerations 

Research basis: No research applicable to this rehabilitation measure has been identified. 

The objective of this rehabilitation method is to enhance the ability of the diaphragm to perform 
adequately and to deliver forces to the vertical elements of the lateral force-resisting system. In 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

Figure 17.4.1-4: Added Collector Anchorage to Shear Wall with 
Cast-In-Place Topping Slab 

order to achieve this, careful consideration of the existing diaphragm configuration and shear 
strength is needed in order to determine what behavior can be achieved and what detailing 
approach is best. In long-span diaphragms, yielding of a chord member may be a preferred 
behavior in order to protect against in-plane shear failure. Where this is true, it should be 
anticipated that the chord member is likely to elongate in one or more locations under tension 
loading, opening up gaps between adjacent precast members. When loading reverses, 
compression will be carried by the chord member until gaps close. In order to prevent local 
buckling failure of the chord, it is advisable to either use a structural steel section that can be 
adequately braced against buckling with a reasonable adhesive anchor spacing, or provide 
confinement of reinforcing bars, as would be provided for a concrete column. These alternatives 
are shown at the right hand side of Figure 17.4.1-1. Both of these alternatives will be more costly 
and difficult to install. In diaphragm configurations where yielding of the chord and collector 
members can be avoided, providing additional reinforcing may be less costly than detailing for 
buckling restraint. Where this approach is taken, a careful evaluation of anticipated forces is 
needed. In general, it is the philosophy of ASCE 7 seismic provisions for inelastic behavior to be 
focused in vertical elements rather than collectors, allowing the vertical elements of the seismic 
force-resisting system to control building dynamic behavior. This may not be achievable in 
diaphragms where the collector serves as a chord for loading in the perpendicular direction, and 
avoiding shear failure is paramount. In this case, use of a detail that restrains buckling is 
recommended. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

Figure 17.4.1-5: Rebar Embedment Considerations in Collector Anchorage 
to Existing Shear Wall 

For the chord configurations shown in Figure 17.4.1-6, the existing original connection (if any) 
likely consists of intermittent cast-in embed plates and a welded plate connection. The capacity 
of anchors embedded in the concrete and welds would have been sized to meet load requirements 
(wind or earthquake) applicable at time of construction. Even if earthquake loading were 
considered, the need to allow for forces in excess of design levels, ductility, and energy 
dissipation likely would not have been considered. Because inelastic behavior in the precast 
walls and diaphragms is very unlikely, it must be anticipated that inelastic behavior will 
concentrate in the connections between members. With many existing connections, evaluation 
would likely identify failure of the anchors embedded in concrete as the weak link. This is an 
undesirable weak link due to lack of ductility. Unless extreme overstrength has been provided, to 
allow the connection to remain elastic, rehabilitation of the connections is needed in order to 
avoid this weak link. ACI 318 Appendix D (ACI, 2005) provisions for anchorage to concrete 
require that that design be governed by tensile or shear strength of a ductile steel element rather 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

the concrete capacity. This requirement is particularly appropriate for rehabilitation of 
connections in precast wall buildings. 

Figure 17.4.1-6: Steel Chord or Collector at Floor Perimeter 
without Cast-In-Place Topping 

Figure 17.4.1-7: Steel Chord or Collector at Floor Perimeter 
without Cast-In-Place Topping 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

The deformation compatibility of the existing and added connections needs to be carefully 
considered. It may be desirable to design new connections to carry all forces, neglecting the 
contribution of the existing connections. In some cases it might even be desirable to cut the weld 
on the original connection to ensure that it will not carry load. 

Detailing Considerations 

Use of reinforcing dowels to anchor new chord or collector reinforcing to existing concrete, as 
shown in Figures 17.4.1-1 and 17.4.1-2, relies on shear friction. Roughening of the existing 
concrete to the amplitude required by ACI 318 requires a very heavy sandblasting or chipping 
with a jackhammer, in order to expose the aggregate. This is expensive and messy. It is often 
preferable to use a reduced  coefficient of less than 1.0 and add more dowels, rather than 
roughening the surface. Removal of finishes and cleaning of the concrete surface is still required. 
Use of shear friction also requires that the yield capacity of the dowel be developed on both sides 
of the joint. This forces use of smaller dowels and a curb dimension adequate to develop 
standard hooks per ACI 318 requirements. Adhesive anchor embedment requirements for 
development of the bar yield are generally available from the manufacturer. 

Figure 17.4.1-2 places new chord or collector reinforcing at the underside of the floor so that 
vehicle or foot traffic above is not disrupted. Where chord or collector reinforcing runs parallel 
to the T-beams, use of reinforcing bars is likely feasible. Where the chord or collector runs 
perpendicular, reinforcing bars would have to be installed in short lengths threaded thru the T-
beam webs. An unstressed tendon that can be more easily placed might be a preferable 
alternative. 

Figure 17.4.1-3 uses the existing girder as the chord or collector member and adds angles for 
shear transfer between the slab and collector. This must be used in combination with enhanced 
connections at the girder ends. The figure illustrates the possible pre-earthquake cracks in the 
topping slab described by Wood, Stanton and Hawkins (2000). The shear transfer angles bypass 
this potentially weak location. Where steel splice angles are used, splice details will be required. 
A field welded splice would be common in concrete rehabilitation. Where yielding of the steel 
chord/collector member can be anticipated, it may be desirable to proportion the splice plates and 
welds to develop the anticipated strength of the chord/collector member. 

Figure 17.4.1-5 illustrates considerations when collector steel is to be doweled into the end of an 
existing shear wall. This is only possible when there is adequate reinforcing within the wall in 
the vicinity of the collector anchorage to distribute the collector force over the wall length. 
Anchoring two No. 9 bars to the end of walls with only two No. 4 bars should not occur. Further, 
it is important that the collector lap adequately with reinforcing in the shear wall that can 
distribute the collector force over the wall length, as shown in the upper two figures. If 
embedment were limited to the requirements of the anchor manufacturer, as shown in the bottom 
figure, failure of the collector may occur due to inability of the wall reinforcement to develop. 

Cost/Disruption and Construction Considerations 

The work required to add or enhancement chords and collectors is generally spread out over 
considerably over the building area. This distribution of work is reasonably easy in parking 
garages because there are generally no finishes to remove and replace, all areas are reasonably 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

accessible for materials and equipment, and user relocation does not involve a big effort. In 
contrast, the distribution of work will create significantly greater cost and coordination in other 
building types such as commercial or residential. In these building types, it may well be easier 
and more cost effective to add vertical elements, rather than enhancing chords and collectors, 
because this work may be located in one or more local areas. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique other than the use of 
proprietary anchors as part of the assemblage. 

17.4.2 Enhance Connections Between Existing Precast Diaphragm, Shear Wall 
and Foundation Elements 
Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses rehabilitation of inadequate connections between precast 
concrete elements including: diaphragm to shear wall, shear wall above to shear wall below, 
shear wall to foundation, and shear wall panel to panel connections. 

Description of the Rehabilitation Technique 

Existing connections of precast wall and diaphragm elements are generally either welded 
connections between cast-in embed plates, or connections involving mechanical, grouted or 
welded rebar splices. Where precast elements are combined with cast-in-place elements, Figures 
17.4.2-1 and 17.4.2-2 show connections between precast wall panels and foundation and 
diaphragms. The connections depict one possible configuration for existing welded connections 
and rehabilitation approaches for enhancing connection shear capacity. The rehabilitation 
measures involve adding new steel angles and adhesive anchors as required to carry seismic 
forces. Figures 17.4.2-1B shows added anchorage to a hollow core wall section using grout, 
which provides a high connection capacity. Alternately an anchor specifically designed for 
attachment to hollow masonry could be used, resulting in a lower connection capacity. 

Design Considerations 

Research basis: FRP composite connections between wall panels have been tested at the 
University of Utah (Pantelides, Volnyy, Gergeley, and Reaveley, 2003). 

Existing precast shear wall connections most often use cast-in embed plates and welded plate 
connections. The capacity of anchors embedded in the concrete and welds would have been sized 
to meet load requirements (wind or earthquake) applicable at time of construction. Even if 
earthquake loading were considered, the need to allow for forces in excess of design levels, 
ductility and energy dissipation likely would not likely have been considered. Because inelastic 
behavior in the precast walls and diaphragms is very unlikely, it must be anticipated that inelastic 
behavior will concentrate in the connections between members. With many existing connections, 
evaluation would likely identify failure of the anchors embedded in concrete as the weak link. 
This is an undesirable weak link due to lack of ductility. Unless extreme overstrength has been 
provided, allowing the connection to remain elastic, rehabilitation of the connections is needed in 
order to avoid this weak link. ACI 318 Appendix D (ACI, 2005) provisions for anchorage to 
concrete require that the design be governed by tensile or shear strength of a ductile steel element 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

rather the concrete capacity. This requirement is particularly appropriate for rehabilitation of 
connections in precast wall buildings. The deformation compatibility of the existing and added 
connections needs to be carefully considered. It may be desirable to design new connections to 
carry all forces, neglecting the contribution of the existing connections. In some cases, it might 
even be desirable to cut the weld on the original connection to ensure that it will not carry load. 

Figure 17.4.2-1: Added or Enhanced Precast Wall-to-Foundation Connection 

Each precast wall panel tends to have two existing welded connections at the wall top and two at 
the wall bottom, and connections are generally not provided panel to panel. As a result, the two 
connections at the wall bottom must resist both shear and overturning forces from in-plane loads 
as well as out-of-plane loads. This is illustrated in Figure 17.4.2-3A. Where multiple wall panels 
are in line, as shown in Figure 17.4.2-3B, it may be possible to add panel to panel connections to 
resist overturning, reducing the demand on the bottom of panel connections. Overturning 
capacity will still be needed at each end of the panel group. The perpendicular wall panel at the 
right side of Figure 17.4.2-3 should not be counted on to resist the uplift, since it may also be 
overturning. 

Figure 17.4.2-4 illustrates an FRP composite connection between wall panels developed and 
tested at the University of Utah (Pantelides, Volnyy, Gergeley, and Reaveley, 2003) that would 
be applicable to this use. While various surface preparations and FRP applications were 
investigated, the researchers settled on use of a high-pressure water jet preparation of the 
concrete surface to expose aggregate, a bonding agent and dry lay-up of carbon fibers, and 
saturation with an epoxy resin. A lay up of six layers of 12k (12,000 threads per tow) carbon 
fiber reinforcing in a 16 inch by 48 inch rectangle on one face of the wall panels provided failure 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

loads on the order of 40 kips. The failures were sudden and brittle, indicating that design of this 
connection type should be considered force-controlled, and ductility should be provided in other 
connections or members. See Section 13.4.1 for general information on FRP composite 
overlays. 

Figure 17.4.2-2: Precast Wall Connection 

The new connections shown have both vertical and horizontal eccentricities that must be 
considered in design of the angle section and the anchors. 

Detailing Considerations 

Anchorage to hollow core wall panels is more difficult than solid precast panels because the 
voids do not leave solid concrete sections large enough to meet embedment and edge distance 
requirements for adhesive anchors. For most anchorage, drilling grout access and inspection 
holes and filling the lower portion of the void with grout can provide a solid cell for anchor 
placement. Similar grouting of floor planks is sometimes used in new construction. For very light 
anchorage loads, it may be possible to use screen-tube adhesive anchors specifically designed for 
attachment to hollow masonry units. Along with low capacity, failure of this anchorage should 
be expected to have brittle behavior. 

17-20 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

Figure 17.4.2-3: Modification of Demand on Anchors 
Through Use of Panel-to-Panel Connections 

Cost, Disruption and Construction Considerations 

The cost of retrofitting connections in areas covered by finishes can be very expensive due to 
finish removal and replacement. The use of high-pressure water jets for surface preparation is not 
practical for an occupied building with finishes. It may be possible in an open building such as a 
parking garage. 

Proprietary Concerns 

FRP systems are proprietary. Manufacturers should be contacted for appropriate uses and 
limitations. Proprietary adhesives are used as part of connection details. 

17-21 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

Figure 17.4.2-4: Panel Setup for Testing of FRP Panel-to-Panel Connections 

17.4.3 Mitigate Configurations Creating Short Columns 
Deficiency Addressed by Rehabilitation Technique 

This technique addresses mitigation of unintended shortening and fixity of concrete columns. 

Description of the Rehabilitation Technique 

Figures 17.4.3-1, 17.4.3-2 and 17.4.3-3 illustrate unintended shortening and stiffening of 
columns, common in parking garages. The condition in Figure 17.4.3-1 is very easily mitigated 
by sawcutting the required gap between the column and the guardrail. The connection at the base 
of the guardrail may need to be improved as a result. The condition in Figure 17.4.3-2 can be 
improved by creating a hinge or joint in the column at the top of slab level. This will require 
shoring to take the load off of the column during modification, and the development of a shear 
and tension connection to the pedestal and footing. The condition in Figure 17.4.3-3 is difficult 
to address. Columns with this configuration are not easily confined. A solution that can minimize 
column shortening for new construction is to provide separate columns for the ramp and level 
floors. This approach could also be attempted for a retrofit, but would be costly. 

Design Considerations 

Research basis: No research applicable to this rehabilitation measure has been identified. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique. 

17-22 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

Figure 17.4.3-1: Accidental Reduction of Column Height Due to Guardrail 

Figure 17.4.3-2: Accidental Increase in Column Fixity 
Due to Embedment into Grade 

17-23 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

Figure 17.4.3-3: Accidental Reduction of Column Height at Ramp 

17.5 References 
ACI, 2005, Building Code Requirements for Structural Concrete (ACI 318), American Concrete 
Institute, Farmington Hills, MI. 

City of Los Angeles and SEAOSC (Structural Engineers Association of Southern California), 

Findings & Recommendations of the City of Los Angeles/SEAOSC Task Force on the Northridge 
Earthquake, 1994, City of Los Angeles, CA. 

EERI, 1989, Armenia Earthquake Reconnaissance Report, Earthquake Spectra Special Edition, 
Earthquake Engineering Research Institute, Oakland, CA, August. 

EERI, 1995, Guam Earthquake of August 8, 1993 Reconnaissance Report, Earthquake Spectra, 
Supplement B to Volume 11, Earthquake Engineering Research Institute, Oakland, CA, April. 

EERI, January 1996, Northridge Earthquake of January 17, 1994, Reconnaissance Report, 
Volume 2, Earthquake Spectra, Supplement C to Volume 11, Earthquake Engineering Research 
Institute, Oakland, CA. 

FEMA, 2004, NEHRP Recommended Provisions for Seismic Provisions for New Buildings and 
Other Structures, FEMA 450, Federal Emergency Management Agency, Washington, D.C. 

17-24 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 17  Type PC2: Precast Concrete Frames with Shear Walls 

Ghosh, S.K., Observations on the Performance of Structures in the Kobe Earthquake of January 
17, 1995, PCI Journal, Vol. 40, No. 2, March/April 1995, Precast/Prestressed Concrete 
Institute, Chicago, IL. 

Iverson, J.K. and N.M. Hawkins, 1994, Performance of Precast/ Prestressed Concrete Building 
Structures During the Northridge Earthquake, PCI Journal, Vol. 39, No. 2, Precast / Prestressed 
Concrete Institute, Chicago, IL. 

Lyons, S., R. Bligh, J. Purlinton, and D. Beaudoin, 2003, What about the Ramp? Seismic 
Analysis and Design of Ramped Parking Garages, Proceedings of the 2003 SEAOC Convention, 
Structural Engineers Association of California, Sacramento, CA. 

Mooradian, D., 2005, personal communication. 

Naito, C. and L. Cao, 2004, Precast Diaphragm Panel Joint Connector Performance, 
Proceedings of the 13th World Conference on Earthquake Engineering, Vancouver, B.C. Canada. 

Nakaki, S., 1998, Design Guidelines: Precast and Cast-in-Place Concrete Diaphragms, 
Earthquake Engineeering Research Institute, Oakland, CA. 

Pantelides, C. P., V. A. Volnyy, J. Gergeley, and L.D. Reavely, 2003, Seismic Retrofit of 
Precast Concrete Panel Connections with Carbon Fiber Reinforced Polymer Composites, PCI 
Journal, Vol. 48, No. 1, January/February, Precast/Prestressed Concrete Institute, Chicago, IL. 

Priestley, M.J.N., S. Sritharan, JR. Conley, and S. Pampanin, 1999, Preliminary Results and 
Conclusions From the PRESSS Five-Story Precast Concrete Test Building, PCI Journal, Vol. 
44, No. 6, November/December, Precast/Prestressed Concrete Institute, Chicago, IL. 

Wan, G., R. Fleischman, J. Restrepo, R. Sause, C. Naito, S.K. Ghosh, L. Cao, and M. Schoettler, 
2004, Integrated Analytical and Experimental Research Program to Develop a Seismic Design 
Methodology for Precast Diaphragms, Proceedings of the 2004 SEAOC Convention, Structural 
Engineers Association of California, Sacramento, CA. 

Wood, S. L., J.F. Stanton, and N.M. Hawkins, 2000, New Seismic Design Provisions for 
Diaphragms in Precast Concrete Parking Structures, PCI Journal, Vol. 45, No. 1, 
January/February 2000, Precast/Prestressed Concrete Institute, Chicago, IL. 

17-25 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 18  Type RM1t: Reinforced Masonry Bearing Walls 

Chapter 18 - Building Type RM1t: Reinforced Masonry 
Bearing Walls (Similar to Tilt-up Concrete Shear Walls) 

18.1 Description of the Model Building Type 
Building Type RM1 is constructed with reinforced masonry (brick cavity wall or concrete 
masonry unit) perimeter walls with a wood or metal deck flexible diaphragm. For this document, 
Building Type RM1 is separated into two categories. Chapter 19 describes RM1u, which is 
multistory, and typically has interior CMU walls and shorter diaphragm spans. This chapter 
covers RM1t, the large, typically one-story buildings with relatively open interiors that are 
similar to concrete tilt-ups. The exterior walls are commonly bearing, with an interior post and 
beam system of steel or wood. Older buildings of this type are generally small and used for a 
wide variety of occupancies. Recently, the building type has become commonly used for one-
story warehouse and wholesale/retail occupancies similar to tilt-up (Building Type PC1) 
buildings. Figure 18.1-1 illustrates one example of the RM1t Building Type. 

Figure 18.1-1: Building Type RM1t: Reinforced Masonry Bearing Walls 

18-1 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 18  Type RM1t: Reinforced Masonry Bearing Walls 

Guidelines for Seismic Evaluation and Rehabilitation of Tilt-Up Buildings and Other Rigid 
Wall/Flexible Diaphragm Structures or SEAONC Guidelines (SEAONC, 2001) provides a 
substantial collection of information on flexible diaphragm / rigid wall building issues, the West 
Coast experience with earthquake performance, rehabilitation priorities, and techniques for 
rehabilitation. This document was a primary source of information for Chapter 16 (PC1 
buildings), and it is also recommended for RM1t buildings. 

Walls 

Exterior reinforced masonry walls are the primary vertical elements in the lateral load-resisting 
system. Buildings with large plan areas may have interior walls providing additional lateral 
resistance; however, this is not common. Like tilt-up construction, masonry walls have recently 
transitioned from use of code-prescribed height to thickness (h/t) limits to use of much higher h/t 
ratios, in combination with rational analysis of slenderness effects. Most existing construction, 
however, will have been designed using prescriptive ratios and allowable stress design methods. 

Reinforced masonry walls in RM1t buildings share many issues with PC1 buildings, discussed 
in Chapter 16. Two distinctive aspects of reinforced masonry walls require discussion: 
movement control joints and partially grouted masonry. 

The masonry industry recommends providing vertical control joints in new masonry wall 
construction to accommodate masonry wall shrinkage and thermal movement. The currently 
recommended maximum spacing of control joints is the lesser of 1.5 times the wall height or 25 
feet (CMACN, 2003). The inclusion of and spacing of control joints varies significantly, 
however, with the age and region of the building. Wall-to-diaphragm anchorage remains a 
primary concern irrespective of whether control joints are provided. Where vertical control joints 
occur, diaphragm chords and collectors will be provided by either horizontal reinforcing that 
continues across the control joint (typically provided at the diaphragm level only), or a steel 
angle or similar member on the face of the masonry. Rehabilitation of chords, collectors, and 
shear transfer for RM1t building will be much the same as PC1 buildings. Masonry wall 
construction without control joints will not have to rely as much on discreet chords and 
collectors; adequacy of shear transfer may still be a focus of rehabilitation. 

In areas of high seismic hazard, it is most common for masonry walls to be fully grouted. In 
many other areas, however, grout is only provided at required reinforcing. A typical reinforced 
masonry wall in a RM1t building might have vertical reinforcing and grout alongside window 
and door openings, and at between four and ten feet on center horizontally and vertically in the 
piers and spandrels. Partial grouting has a significant effect on the weight of the wall and 
calculated seismic forces, as well as wall strength for both in-plan and out-of-plane forces. 
Where partial grouting has been provided, grout locations need to be known in order to design 
wall to diaphragm anchorage. 

Gravity Load-Carrying Support at the Building Perimeter 

It is most common for wood girders to be supported on the building exterior walls in RM1t 
buildings. As in PC1 buildings, girder connections to the exterior walls are required to resist wall 
out-of-plane loading in addition to gravity loads. Connections of girders to the exterior walls 
require evaluation and possible rehabilitation. Section 16.4.2 discusses applicable rehabilitation 

18-2 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 18  Type RM1t: Reinforced Masonry Bearing Walls 

measures. Where the existing masonry is partially grouted, it may be necessary to open up the 
masonry face and grout at new anchorage locations. Where this is done, cast-in anchors can be 
provided in lieu of adhesive anchors. 

Roof Diaphragms 

Like the PC1 building, the roof system will generally be either of wood or steel construction. In 
the western states, roof systems are almost exclusively wood sheathed. Outside of the western 
states, roof systems are almost exclusively sheathed with steel decking topped with rigid 
insulation or vermiculite concrete. For both the wood and the steel roof systems, the roof 
diaphragm in the RM1t building is almost always flexible compared to the walls. See Chapter 16 
for additional discussion. 

Wall-to-Diaphragm Connections 

Like PC1 buildings, wall-to-diaphragm connections are thought to be the aspect of RM1t 
buildings most vulnerable to earthquake damage, due to the significant force and deformation 
demands imposed on this connection. The wall-to-diaphragm connections are therefore 
recommended as the first focus of rehabilitation measures for one-story RM1t buildings. See 
Chapter 16 for a discussion of past performance of these connections in tilt-up buildings. Similar 
to tilt-up buildings, even in RM1t buildings constructed or upgraded recently, it should be 
assumed that these connections require review and possible rehabilitation. 

Interior Additions 

Mezzanines and interior second stories, commonly constructed within large box-like RM1t 
buildings, can restrain building movement under earthquake loading, resulting in unintended 
load paths and damage. See Chapter 16 for discussion. 

Foundations 

RM1t buildings are generally constructed on continuous perimeter footings, with dowels to the 
reinforced masonry walls at vertical reinforcing locations. 

18.2 Seismic Response Characteristics 
RM1t buildings, like PC1 buildings, are distinguished by rigid shear walls and flexible 
diaphragms. Like PC1 buildings, amplification of seismic forces near the center of the 
diaphragm is of concern for both diaphragm capacity and wall anchorage capacity. The in-place 
construction of the reinforced masonry walls may provide wall continuity that makes RM1t 
buildings somewhat less vulnerable to partial collapse than PC1 buildings; however, the 
potential for significant performance problems exists. 

18.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
Unlike PC1 buildings, damage to RM1t buildings has not been noted as significant or wide 
spread. In the 1994 Northridge earthquake, little in the way of damage to RM1t buildings was 
reported (EERI, 1996; and Klingner, 1994). One item of interest was damage to masonry walls at 
building corners near the roof line, attributed to interaction between the masonry wall and 
flexible wood diaphragm. See below for general discussion and Table 18.3-1 for a detailed 

18-3 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 18  Type RM1t: Reinforced Masonry Bearing Walls 

Table 18.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM1t Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Global 
Strength 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [21.4.8] 
Concrete wall overlay 
[21.4.5] 
Infill openings 
Global 
Stiffness 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [21.4.2] 
Concrete wall overlay 
[21.4.5] 
Infill openings 
Configuration Torsionally irregular 
plans 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [21.4.8] 
Concrete wall overlay 
[21.4.5] 
Infill openings 
Re-entrant corners Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [21.4.8] 
Collector [7.4.2] 
Enhance existing 
collector 
Concrete wall overlay 
[21.4.5] 
Infill openings 
Collector [7.4.2] 
Incidental bracing Separate 
component from 
incidental bracing 
Load Path Inadequate or 
missing wall-todiaphragm 
tie for 
out-of-plane load 
Wall-to-diaphragm 
tension anchors plus 
subdiaphragms and 
cross-tie [16.4.1] 
Inadequate 
anchorage to 
diaphragms for in-
plane forces 
Wall-to-diaphragm 
shear anchors 
[21.4.2] 
Inadequate 
collectors 
Add collector [7.4.2] Improve collector 
member and 
connections 

18-4 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 18  Type RM1t: Reinforced Masonry Bearing Walls 

Table 18.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM1t Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Component 
Detailing 
Wall inadequate for 
out-of-plane bending 
Wall strongback or 
pilaster [21.4.3] 
Inadequate detailing 
of slender walls 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [21.4.8] 
Concrete wall overlay 
[21.4.5] 
Infill openings 
Add backup vertical 
supports where 
bearing might be lost 
[21.4.11] 
Diaphragms Inadequate in-plane 
strength and/or 
stiffness 
Steel braced frame 
[7.4.1] 
Concrete/masonry 
shear wall [21.4.8] 
Enhance existing 
diaphragm [22.2.1] 
Horizontal braced 
frame [21.2.10] 
Inadequate chord 
capacity 
Enhance chord 
[22.2.2] 
Enhance chord 
[22.2.2] 
Excessive stresses at 
openings and 
irregularities 
Enhance diaphragm 
detailing 
Re-entrant corners Enhance diaphragm 
detailing 
Foundations See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation measures. 

18-5 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 18  Type RM1t: Reinforced Masonry Bearing Walls 

compilation of common seismic deficiencies and rehabilitation techniques for Building Type 
RM1t. See also Chapter 16 for similar issues in PC1 buildings. 

Global Strength and Stiffness 

Global strength and stiffness are rarely a concern for large box-like RM1t buildings, but can be 
for smaller buildings that have very short walls along a street-front side. Rehabilitation of global 
strength and stiffness deficiencies will commonly involve adding new vertical elements, 
enhancing existing elements, or infilling openings in existing walls. 

Configuration 

Poor distribution of shear walls can result in torsionally irregular behavior of RM1t buildings. 
Common occurrences include street-front walls in commercial buildings. The most direct 
approach to rehabilitation of this condition is the addition of strength and stiffness in line with 
the perforated wall. This can be accomplished through addition of new shear walls, enhancing 
existing shear walls, or addition of steel braced frames. 

Rehabilitation at re-entrant corners requires the provision of adequate chords and collectors, 
shear transfer to the in-set wall panels, and possibly the strengthening of the wall panels and 
connections to the foundation. The SEAONC Guidelines suggest that there may be diaphragm 
continuity over this interior diaphragm support, increasing the diaphragm reaction to the in-set 
wall line. See Chapter 16 for illustration and additional discussion. 

Load Path 

As previously mentioned, load path connections between the masonry walls and the flexible 
diaphragm are suggested as the first focus of rehabilitation in RM1t buildings. Diaphragm cross-
ties, as addressed in Chapter 22, are a required continuation of the wall anchorage system. 
Section 16.4.1 discusses applicable rehabilitation measures. Where the existing masonry is 
partially grouted, it may be necessary to open up the masonry face and grout at new anchorage 
locations. Where this is done, cast-in anchors can be provided in lieu of adhesive anchors. 
Connection between the wall and diaphragm may also be inadequate for in-plane shear loads. 

The addition of or enhancement of existing collectors may be required in order to transmit 
diaphragm forces to the resisting shear walls. This is particularly of concern when a limited 
length of shear wall intended to carry a significant portion of the building shear. Although not as 
common, there is also significant concern when vertical offsets in the roof diaphragm result in 
incomplete chords or collectors. Any breaks or offsets in chords or collectors need to be carefully 
evaluated. 

Component Detailing 

Component detailing deficiencies include inadequate out-of-plane wall capacity. Where existing 
walls are partially grouted concrete masonry, it may be possible to place additional vertical 
reinforcing and grout in ungrouted cells, accessed by cutting open face shells. Where this 
approach is taken, doweling of the vertically reinforcing at the top and bottom of the grouted 
cells would commonly be provided. In addition, the increase in wall weight should be considered 
in building seismic forces. Where existing masonry wall construction is solid, it is seldom 
practical to address wall capacity by adding reinforcing and concrete thickness to individual wall 

18-6 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 18  Type RM1t: Reinforced Masonry Bearing Walls 

sections, so addition of wall pilasters or strongbacks is common. Where pilasters are added to 
masonry walls, pilaster-to-roof diaphragm anchorage must be provided to accommodate the 
concentration of wall out-of-plane force. 

Diaphragm Deficiencies 

Due to changes in building code requirements, it is very common for diaphragms in areas of high 
seismic hazard to have inadequate in-plane shear capacity. These diaphragms may also have 
inadequate in-plane stiffness due to high unit shear stresses. Regardless of this, the SEAONC 
Guidelines indicate that diaphragm overstresses have rarely been associated with significant 
earthquake damage. Diaphragm strength and stiffness deficiencies are most often rehabilitated by 
enhancing the existing diaphragm. 

Other diaphragm deficiencies include inadequate chord capacity and stress concentrations at 
large diaphragm openings and re-entrant corners. Rehabilitation at re-entrant corners primarily 
involves the provision of adequate chords and collectors. The same is true at large diaphragm 
openings. 

18.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
No techniques have been developed for this building type. See other chapters for detailed 
descriptions of relevant rehabilitation techniques. 

18.5 References 
CMACN, 2003, Movement Control Joints, Masonry Chronicles Winter 02-03, Concrete 
Masonry Association of California and Nevada, Citrus Heights, CA. 

EERI, January 1996, Northridge Earthquake of January 17, 1994, Reconnaissance Report, 
Volume 2, Earthquake Spectra, Supplement C to Volume 11, Earthquake Engineering Research 
Institute, Oakland, CA. 

Klingner, R., 1994, Performance of Masonry Structures in the Northridge California Earthquake 
of January 17, 1994 (Technical Report 301-94), The Masonry Society, Boulder, CO. 

SEAONC (Structural Engineers Association of Northern California), 2001, Guidelines for 
Seismic Evaluation and Rehabilitation of Tilt-Up Buildings and Other Rigid Wall/Flexible 
Diaphragm Structures, Structural Engineers Association of Northern California, San Francisco, 
CA. 

18-7 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 19  Type RM1u: Reinforced Masonry Bearing Walls 

Chapter 19 - Building Type RM1u: Reinforced Masonry 
Bearing Walls (Similar to Unreinforced Masonry Bearing 
Walls) 

19.1 Description of the Model Building Type 
Building Type RM1 takes a variety of configurations, but they are characterized by reinforced 
masonry walls with flexible diaphragms such as wood or metal deck. The walls are commonly 
bearing, but the gravity system also contains post and beam construction of wood or steel in 
interior or some faade locations. For this document, Building Type RM1 is separated into two 
categories. Chapter 18 describes RM1t, the large, typically one-story buildings with relatively 
open interiors that are similar to concrete tilt-ups. This chapter covers RM1u, which is 
multistory, and typically has interior CMU walls and shorter diaphragm spans. It is similar to 
Building Type URM (Chapter 21) and has many of the same deficiencies. Figure 19.1-1 shows 
an example of this building type. 

Figure 19.1-1: RM1u Building Type: Reinforced Masonry Bearing Walls 

19-1 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 19  Type RM1u: Reinforced Masonry Bearing Walls 

Masonry Wall Materials 

FEMA 306 (FEMA, 1999) identifies several common reinforced masonry wall types. These are: 

Fully-grouted hollow concrete block 

Partially-grouted hollow concrete block 

Fully-grouted hollow clay brick 

Partially-grouted hollow clay crick 

Grouted-cavity wall masonry (two wythes of clay brick or hollow units with a reinforced 

grouted cavity) 

Brick veneer facing may be placed on the exterior faade with the above walls used as backing 
walls. 

Floor and Roof Diaphragm 

Floor and roof diaphragm construction is similar to those of Building Type URM, although 
unfilled metal deck diaphragms can be found at the roof and occasionally at floors. See Chapter 

21. 
Foundations 

Foundations for Building Type RM1u are typically spread footings at interior columns and strip 
footings under masonry bearing walls. Footings are typically concrete. 

19.2 Seismic Response Characteristics 
As a flexible diaphragm, stiff wall structure Building Type RM1u is expected to have dynamic 
behavior similar to that described for Building Type URM. See Chapter 21. Since the walls are 
reinforced, however, in-plane and out-of-plane wall behavior modes are substantially different 
from those of unreinforced masonry walls. They are instead more similar to those of reinforced 
concrete. FEMA 306 provides the most comprehensive categorization of reinforced masonry 
wall behavior modes. 

19.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
RM1u buildings, while similar to URM buildings, are generally considered to be less hazardous. 
In-plane damage to reinforced masonry walls is much less likely to reach levels compromising 
life safety. Parapets can still be overstressed, but the risk to life safety is much less than those of 
unreinforced parapets, and out-of-plane failures of the walls spanning between diaphragms are 
relatively unlikely. The most significant risk to loss of life is due to inadequate connections 
between the walls and diaphragms. See below for general discussion and Table 19.3-1 for a 
detailed compilation of common seismic deficiencies and rehabilitation techniques for Building 
Type RM1u. 

Global Strength 

As shear wall buildings, global strength in RM1u buildings is dependent on the in-plane shear 
capacity of the walls. Relatively large seismic forces are needed to lead to life safety concerns 

19-2 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 19  Type RM1u: Reinforced Masonry Bearing Walls 

Table 19.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM1u Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Global 
Strength 
Insufficient in-
plane wall 
strength 
Wood structural panel 
shear wall [5.4.1], [6.4.2] 
Concrete/masonry shear 
wall [21.4.8] 
Steel braced frame [7.4.1] 
Steel moment frame 
[21.4.9] 
Concrete wall overlay 
[21.4.5] 
Fiber composite wall 
overlay [21.4.6] 
Grouting 
Infill openings 
[21.4.7] 
Seismic isolation 
[24.3] 
Global 
Stiffness 
Configuration Soft story, weak 
story, excessive 
torsion 
Wood structural panel 
shear wall [6.4.2] 
Concrete/masonry shear 
wall [21.4.8] 
Steel braced frame [7.4.1] 
Steel moment frame 
[21.4.9] 
Load Path Inadequate or 
missing wall-todiaphragm 
tie 
Tension anchors 
[16.4.1] 
Shear anchors 
[21.4.2] 
Subdiaphragms and 
cross-ties [22.2.3] 
Missing collector Add collector [7.4.2] 
Inadequate girder-
to-column 
connection 
Improve connection 
Supplemental vertical 
supports [21.4.11] 
Inadequate wall-
foundation dowels 
Wall-to-foundation 
improvements 

19-3 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 19  Type RM1u: Reinforced Masonry Bearing Walls 

Table 19.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM1u Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Component 
Detailing 
Wall inadequate 
for out-of-plane 
Exposed interfloor 
wall supports [21.4.3] 
Reinforced cores 
[21.4.4] 
Concrete wall 
overlays [21.4.5] 
Fiber composite 
overlays [21.4.6] 
Poorly anchored 
veneer or 
appendages 
Add ties Remove veneer 
or appendages 
Diaphragms Inadequate 
in-plane strength 
and/or stiffness 
Wood structural panel 
shear wall [6.4.2] 
Steel braced frame 
[7.4.1] 
Concrete/masonry shear 
wall [21.4.8] 
Add wood structural 
panel or moment frame 
crosswall [21.4.10] 
Horizontal braced frame 
[22.2.9] 
Enhance existing 
diaphragm [22.2.1] 
Enhance crosswall 
[21.4.10] 
Inadequate chord 
capacity 
Add steel strap or angle 
Excessive stresses 
at openings and 
irregularities 
Add wood or steel strap 
reinforcement 
Re-entrant corner Wood structural panel 
shear wall [6.4.2] 
Concrete/masonry shear 
wall [21.4.8] 
Steel braced frame [7.4.1] 
Steel moment frame 
[21.4.9] 
Collector [7.4.2] 
Foundations See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 

19-4 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 19  Type RM1u: Reinforced Masonry Bearing Walls 

with in-plane wall behavior, though cracking damage will occur in relatively moderate events. 
When walls are found to be deficient, new vertical lateral force-resisting elements can be added 
at interior locations or existing walls can be enhanced. 

Global Stiffness 

Reinforced masonry walls are generally quite rigid, even if punctured with window openings, so 
global stiffness deficiencies are relatively uncommon. 

Configuration 

Many commercial RM1u buildings will have a fairly open street faade at the ground level, 
leading to a weak and soft first story and torsional irregularities. This is usually addressed by the 
addition of a moment frame at the faade or another vertical lateral force-resisting element at 
some distance back from the faade. 

Load Path 

As noted above, it is the lack of adequate ties between the walls and diaphragms that is the single 
most significant deficiency in RM1u buildings. Rehabilitation measures include tension ties for 
out-of-plane forces and shear ties for in-plane forces along the typical wall-diaphragm interface. 
In many reinforced masonry walls, pilasters are formed by thickening the wall in order to 
provide support of key girder lines. They may not have adequate bearing length for the girder 
seat or sufficient reinforcement at the top of the pilaster. Anchor bolts from the ledger to the 
wall are likely to be present, but cross-grain bending under out-of-plane tension loading will be a 
common deficiency. Rehabilitation measures include supplemental column supports and 
connection enhancements. 

Component Detailing 

When brick veneer is present, it may not be adequately anchored back to the backing masonry, 
creating a falling hazard. Veneer ties can be added. 

Diaphragm Deficiencies 

Wood diaphragms may lack both strength and stiffness in RM1u buildings. This can be 
addressed by adding new interior elements to cut the diaphragm span or by enhancing the 
diaphragm itself with wood structural panel overlays. 

Foundation Deficiencies 

Foundation deficiencies for existing elements are relatively uncommon in RM1u buildings. 
Foundation rehabilitation work usually is focused on the support for new lateral force-resisting 
elements that are added to the superstructure. 

19.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
Rehabilitation techniques for Building Type RM1u are typically similar to those used in URM 
buildings. See Chapter 21 for detailed descriptions of techniques. When metal deck floors are 
present, techniques in Chapter 16 provide examples of rehabilitation methods for connecting the 
metal deck to the tilt-up concrete walls; details for connecting to reinforced masonry walls are 
similar. 

19-5 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 19  Type RM1u: Reinforced Masonry Bearing Walls 

19.5 References 
FEMA, 1999, Evaluation of Earthquake Damaged Concrete and Masonry Wall Buildings: Basic 
Procedures Manual, FEMA 306, May. 

19-6 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 20  Type RM2: Reinforced Masonry Bearing Walls 

Chapter 20 - Building Type RM2: Reinforced Masonry 
Bearing Walls (Similar to Concrete Shear Walls with Bearing 
Walls) 

20.1 Description of the Model Building Type 
This building consists of reinforced masonry walls and concrete slab floors that may be either 
cast-in-place or precast. In this type of building, all walls usually act as both bearing and shear 
walls. The building type is similar and often used in the same occupancies as Building Type 
C2b, namely in mid- and low-rise hotels and motels. This system is also commonly used in 
residential apartment/condominium type buildings. 

Figure 20.1-1: RM2 Building Type: Reinforced Masonry Bearing Walls 

Variations Within the Building Type 

In order for this framing system to be efficient, a regular and repeating pattern of bearing walls 
are required to provide support points for the floor framing. In addition, since it is difficult and 
expensive to make significant changes in the plan during the life of the building, planning 
flexibility is not normally an important characteristic when this structural system is employed. 
The occupancy types that most often fit these characteristics are residential buildings, including 

20-1 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 20  Type RM2: Reinforced Masonry Bearing Walls 

dormitories, apartments, motels, and hotels. These buildings will often be configured with 
reinforced masonry bearing walls between rooms which act as shear walls in the transverse 
direction, and reinforced masonry walls on the interior corridor which act as shear walls in the 
longitudinal direction. Sometimes the longitudinal lateral system includes the exterior wall 
system, although this wall is normally made as open as possible. In any case, the wide variation 
in structural layouts and occupancies that is included in other shear wall buildings such as C2f is 
not seen in RM2. 

It is seldom possible to plan a building layout that provides complete gravity support with walls, 
and often local areas are supported with isolated columns, sometimes with beams and girders. 
However, story heights in these buildings are usually small, and added depth in the floor framing 
system is difficult to obtain. The extent of such beam and column framing often causes 
confusion as to the classification of the structure as a bearing wall system. However, if 
significant plan area is supported solely by walls, the structures are normally classified as RM2. 

There are important variations in floor framing systems employed in this building type, and their 
adequacy to act as a diaphragm is an important characteristic of this building type as discussed 
below. 

Floor and Roof Diaphragms 

The parallel layouts of supporting walls and the need to minimize story heights normally leads to 
the use of one-way uniform-depth concrete floor systems. Cast-in-place and precast systems, 
both conventionally reinforced and prestressed, have been employed. The precast systems are 
often built up of narrow planks, which may not provide an adequate diaphragm unless a cast-inplace 
topping is provided. In addition, the precast systems may be placed with only a very 
narrow bearing area on the supporting walls, almost always on the outer masonry wythe or the 
CMU shell. When prestressed, the planks may be connected into the wall system only with the 
tail of the stressing tendon, and this connection may be inadequate to provide vertical support 
during seismic movements. The adequacy of the shear connection between slab and walls is also 
often an issue for both cast-in-place and precast systems. 

Foundations 

The bearing walls obviously require some kind of starter beam at grade for construction 
purposes, and this often leads to a simple continuous grade beam system. In poor soils, piles or 
drilled piers may be added below the grade beam. A continuous mat foundation may also be 
employed due to the short spans and total length of bearing points in this building type. 

20.2 Seismic Response Characteristics 
Due to the extent of wall, bearing wall buildings will be quite stiff. Elastic and early post-elastic 
response will therefore be characterized with lower-than average drifts and higher-than-average 
floor accelerations. Damage in this range of response should be minimal. 

Overall post-elastic response may often include rocking at the foundation level. If rocking does 
not occur, the height-to-length ratio of shear walls in these buildings may force shear yielding 
near the base, which may lead to strength and stiffness degradation. 

20-2 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 20  Type RM2: Reinforced Masonry Bearing Walls 

Global stability may also be compromised by poor connections between floor slab construction 
and bearing walls. 

Shear Wall Behavior 

When subjected to every increasing lateral load, individual shear walls or piers will first often 
force yielding in spandrels, slabs, or other horizontal components restricting their drift, and 
eventually walls and piers either rock on their foundations, suffer shear cracking and yielding, or 
form a flexural hinge near the base. Shear and flexural behavior are quite different, and 
estimates of the controlling action are affected by the distribution of lateral loads over the height 
of the structure. 

Yielding of spandrels, slabs, or other coupling beams can cause a significant loss of stiffness in 
the structure. Flexural yielding will tend to maintain the strength of the system, but shear 
yielding, unless well detailed, will degrade the strength of the coupling component and the 
individual shear wall or pier will begin to act as a cantilever from its base. In this building type, 
the coupling elements are often slabs, and their lack of bending stiffness may reduce or eliminate 
significant coupling action. 

Rocking is often beneficial, limiting the response of the superstructure. However, the amplified 
drift in the superstructure from rocking must be considered. In addition, if varying wall lengths 
or different foundation conditions lead to isolated or sequencing rocking, the transfer of load 
from rocking walls must be investigated. In buildings with basements, the couple created from 
horizontal restraint at the ground floor diaphragm and the basement floor/foundation (often 
termed the backstay effect) may be stiffer and stronger than the rocking restraint at the 
foundation and should be considered in those configurations. 

Shear cracking and yielding of the wall itself are generally considered undesirable, because the 
strength and stiffness will quickly degrade, increasing drifts in general, as well as potentially 
creating a soft story or torsional response. However, in accordance with FEMA 356 (FEMA, 
2000), shear yielding walls or systems can be shown to be adequate for small target 
displacements. Type RM2 buildings will often fall into this category. 

Flexural hinging is considered ductile in FEMA 356 and will degrade the strength of the wall 
only for larger drifts. Similar to rocking, the global effect of the loss of stiffness of a hinging 
wall must be investigated. 

20.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
See Table 20.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 
Selected deficiencies are further discussed below by category. 

Global Strength 

Due to the extensive use of walls, buildings of this type seldom have deficiencies in this 
category, unless significant degradation of strength occurs due to shear failures. 

20-3 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 20  Type RM2: Reinforced Masonry Bearing Walls 

Table 20.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM2 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Global 
Strength 
Insufficient in-
plane wall shear 
strength 
Concrete/masonry 
shear wall [12.4.2] 
Concrete wall overlay 
[21.4.5] 
Fiber composite wall 
overlay [13.4.1] 
Steel wall overlay 
Seismic 
isolation [24.3] 
Reduce flexural 
capacity 
[13.4.4] 
Insufficient 
flexural capacity 
Concrete/masonry 
shear wall [12.4.2] 
Add chords [12.4.3] 
Inadequate 
capacity of 
coupling beams 
Concrete/masonry 
shear wall [12.4.2] 
Strengthen beams 
[13.4.2] 
Improve ductility of 
beams [13.4.2] 
Remove beams 
Global 
Stiffness 
Excess drift 
(normally near the 
top of the 
building) 
Concrete/masonry 
shear wall [12.4.2] 
Concrete/steel column 
jackets [12.4.5] 
Provide detailing of all 
other elements to accept 
drifts 
Concrete wall overlay 
[21.4.5] 
Supplemental 
damping [24.4] 
Configuration Discontinuous 
walls 
Add wall or adequate 
columns beneath 
[12.4.2] 
Fiber composite wrap of 
supporting columns 
[12.4.4] 
Concrete/steel jacket of 
supporting columns 
[12.4.5] 
Improve connection 
to diaphragm 
[13.4.3] 
Remove wall 
Soft story or weak 
story 
Add strength or 
stiffness in story to 
match balance of 
floors 
Re-entrant corner Add floor area to 
minimize effect of 
corner 
Provide chords in 
diaphragm [12.4.3] 

20-4 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 20  Type RM2: Reinforced Masonry Bearing Walls 

Table 20.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM2 Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Configuration 
(continued) 
Torsional layout Add balancing walls 
[12.4.2] 
Load Path Inadequate 
collector 
Add steel or concrete 
collector [12.4.3] 
Inadequate slab 
bearing on walls 
Add diagonal dowels 
[13.4.3] 
Add steel ledger 
[13.4.3] 
Component 
Detailing 
Wall inadequate 
for out-of-plane 
bending 
Add strongbacks 
[21.4.3] 
Concrete wall overlay 
[21.4.5] 
Wall shear critical Concrete wall overlay 
[21.4.5] 
Fiber composite wall 
overlay [13.4.1] 
Reduce flexural 
capacity [13.4.4] 
Diaphragms Precast 
components 
without topping 
Improve interconnection 
[22.2.11] 
Add topping 
Inadequate in-
plane shear 
capacity 
Concrete slab overlay 
Fiber composite overlays 
[22.2.5] 
Inadequate shear 
transfer to walls 
Add diagonal drilled 
dowels [13.4.3] 
Add steel angle ledger 
[13.4.3] 
Inadequate chord 
capacity 
New concrete or steel 
chord member 
[12.4.3] 
Excessive stresses 
at openings and 
irregularities 
Add chords [12.4.3] Infill openings 
[22.2.4] 
Foundation See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 

20-5 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 20  Type RM2: Reinforced Masonry Bearing Walls 

Global Stiffness 

Similar to strength, global stiffness is seldom a problem in this building type. However the 
effect of coupling slabs on initial stiffness and the potential change in stiffness due to yielding of 
these coupling slabs or wall-beams over doors should be investigated. 

Configuration 

The most common configuration deficiencies in this building type are weak or soft stories 
created by walls that change configuration or are eliminated at the lower floors. It is difficult to 
provide the needed ductility at the weak story and often strength must be added. Completely 
discontinuous walls also create a load transfer deficiency for both overturning and shear. In such 
cases, collectors are often needed in the floor diaphragm, and supporting columns need axial 
strengthening. 

Load Path 

A common deficiency in this building is weakness in the load path from floor to walls, either 
collector weaknesses or shear transfer weakness immediately at the floor wall interface. Local 
transfer can be strengthened by adding concrete or steel corbel elements, dowels, or 
combinations of these components. As indicated above, discontinuous walls also often create 
load path deficiencies from the wall into the diaphragm at the discontinuity. 

Component Detailing 

The most common detailing problem in this building type is an imbalance of shear and flexural 
strength in the walls, leading to pre-emptive shear failure. This condition may be shown to be 
acceptable with small displacement demands. Walls can be strengthened in shear with overlays 
of concrete, steel, or FRP. 

The layout of walls often forces coupling between walls through the slab system or across 
headers of vertically aligned doors. These coupling components are seldom designed for the 
coupling distortions that they will undergo, particularly in older buildings. Short lengths of slabs 
between adjacent walls receive damage by coupling action that could compromise the gravity 
capacity. It is difficult to add strength or ductility to these slab areas, but vertical support at 
support points can be supplemented by corbels of steel or concrete. Damage to headers over 
doors often does not contribute to deterioration of overall response and can sometimes be 
acceptable. Local areas of wall can also be strengthened by overlays of concrete, steel, or FRP. 

Diaphragm Deficiencies 

Precast floor systems used in this building type often provide inadequate diaphragm behavior 
that could lead to bearing failures at the floor wall interface, particularly when no topping slab is 
present. Some topping slabs used primarily for leveling and smoothing the floor are 
inadequately tied to the precast elements or the walls, and are too thin or poorly reinforced to act 
as diaphragms on their own. See Chapter 22. 

20-6 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 20  Type RM2: Reinforced Masonry Bearing Walls 

Foundation Deficiencies 

This building type often places large demands on the foundation system. If rocking is shown to 
be a controlling displacement fuse for the building, the foundations must be investigated to 
assure that these displacements can safely occur. See Chapter 23. 

20.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
Most significant recommendations listed in Table 20.3-1 are the same as techniques used in the 
similar building type, C2b, Concrete Shear Walls (Bearing Wall Systems) or general techniques 
applied to concrete diaphragms. Details concerning these techniques can be found in other 
chapters. 

20.5 References 
FEMA, 2000, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 
356, Federal Emergency Management Agency, Washington, D.C., November. 

20-7 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Chapter 21 - Building Type URM: Unreinforced Masonry 
Bearing Walls 

21.1 Description of the Model Building Type 
Building Type URM consists of unreinforced masonry bearing walls, usually at the perimeter 
and usually brick masonry. The floors are typically of wood joists and wood sheathing 
supported on the walls and on interior post and beam construction. This building type is 
common throughout the United States and was built for a wide variety of uses, from one-story 
commercial or industrial occupancies to multistory warehouses to mid-rise hotels. It has 
consistently performed poorly in earthquakes. The most common failure is an outward collapse 
of the exterior walls caused by loss of lateral support due to separation of the walls from the floor 
and roof diaphragms. Figure 21.1-1 shows an example of this building type. 

Building Type URMA is similar to the Building Type URM, but the floors and roof are 
constructed of materials that form a rigid diaphragm, usually concrete slabs or steel joists with 
flat-arched unreinforced masonry spanning between the joists. Building Type URMA is not 
covered by this document. 

Figure 21.1-1: Building Type URM: Unreinforced Masonry Bearing Walls 

Masonry Wall Materials 

FEMA 306 (FEMA, 1999a) provides an overview of masonry wall material variables. It is 
paraphrased here. Unreinforced masonry is one of the oldest and most diverse building 

21-1 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

materials. Important material variables include masonry unit type, wall construction type, and 
material properties of various constituents. 

Solid clay-brick unit masonry is the most common type of masonry unit, but there are a number 
of other common types, such as hollow clay brick, structural clay tile, concrete masonry, stone 
masonry, and adobe. Hollow clay tile (HCT) is a more common term for some types of 
structural clay tile. Concrete masonry units (CMU) can be ungrouted, partially grouted, or fully 
grouted. Stone masonry can be made from any type of stone, but sandstone, limestone, and 
granite are common. Other stones common in a local area are used as well. Sometimes 
materials are combined, such as brick facing over CMU backing, or stone facing over a brick 
backing. 

Wall construction patterns also vary widely, with bond patterns ranging from common running 
bond in brick to random ashlar patterns in stone masonry to stacked bond in CMU buildings. 
The variety of solid brick bond patterns is extensive. Key differences include the extent of 
header courses, whether collar joints are filled, whether cavity-wall construction was used, and 
the nature of the ties between the facing and backing wythes. In the United States, for example, 
typical running-bond brick masonry includes header courses interspersed by about five to six 
stretcher courses. Header courses help tie the wall together and allow it to behave in a more 
monolithic fashion for both in-plane and out-of-plane demands. The 1997 UCBC (ICBO, 1997) 
and 2003 IEBC (ICC, 2003) have specific prescriptive requirements on the percentage, spacing, 
and depth of headers. Facing wythes not meeting these requirements must be considered as 
veneer and are therefore not used to determine the effective thickness of the wall. Veneer 
wythes must be tied back to the backing to help prevent out-of-plane separation and falling 
hazards. Although bed and head joints are routinely filled with mortar, the extent of collar-joint 
fill varies widely. Completely filled collar joints with metal ties between wythes help the wall to 
behave in a more monolithic fashion for out-of-plane demands. One form of construction where 
interior vertical joints are deliberately not filled is cavity-wall construction. Used in many 
northeastern United States buildings, the cavity helps provide an insulating layer and a means of 
dissipating moisture. The cavity, however, reduces the out-of-plane seismic capacity of the wall. 

Material propertiessuch as compressive, tensile and shear strengths and compressive, and 
tensile and shear modulivary widely among masonry units, brick and mortar. An important 
issue for in-plane capacity is the relative strength of masonry and mortar. Older mortars 
typically used a lime/sand mix and are usually weaker than the masonry units. With time, 
cement was added to the mix and mortars became stronger. When mortars are stronger than the 
masonry, strength may be enhanced, but brittle cracking through the masonry units may be more 
likely to occur, resulting in lower deformation capacity. 

Given the wide range of masonry units, construction and material properties, developing 
comprehensive mitigation techniques for all permutations is not practical. The rehabilitation 
measures in this document are most directly relevant to solid clay brick masonry laid in running 
bond with a typical spacing of header courses. 

For additional general background on URM materials, see ABK (1981a), FEMA 274 (FEMA, 
1997b), FEMA 307 (FEMA, 1999b), and Rutherford and Chekene (1997). 

21-2 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Floor and Roof Diaphragms 

Building Type URM, by definition, is built with diaphragms that are considered, relative to the 
masonry walls, to be flexible. Typically, wood sheathing is attached to wood joists. Several 
types of wood diaphragm construction are common, including: 

Roofs with straight sheathing and roofing applied directly to the sheathing 

Roofs with diagonal sheathing and roofing applied directly to the sheathing 

Floors with straight tongue-and-groove flooring 

Floors with straight tongue-and-groove flooring over straight sheathing 

Floors with finished flooring over diagonal sheathing 

Wood structural panel overlays may have been added as part of past renovation work, or there 
may be additional layers of sheathing materials. In some buildings with heavy live loads, like 
warehouses, 2x or 3x decking may have been used to span between joists. 

Foundations 

Foundations for URM buildings typically are spread footings at interior columns and strip 
footings under masonry bearing walls. Footings are typically either brick or concrete, though 
stone might be found under older walls, particularly if stone masonry was used in the walls. 

21.2 Seismic Response Characteristics 
In many building types, the horizontal diaphragms are more rigid than the vertical elements of 
the lateral force-resisting system. Such buildings are often thought of as lumped mass systems 
with the weight tributary to each diaphragm level lumped along a vertical cantilever with 
dynamic properties dependent on the stiffness of the vertical lateral force-resisting elements. 
Ground motion input at the base is dynamically amplified up the cantilever, increasing at each 
floor level. Each point within a floor has a similar acceleration. 

Building Type URM, by contrast, has flexible diaphragms and stiff walls. Beginning with the 
ABK research program in the 1980s (see ABK, 1981a,b,c; and ABK, 1984), a different dynamic 
model was formulated for URM buildings. The ABK model assumes that there is relatively little 
dynamic amplification between the base and the top of the URM walls in the direction parallel to 
input motion. Significant amplification instead occurs at the midspan of the flexible diaphragms 
as they are driven by in-plane motion of the end walls. This generates large out-of-plane forces 
on the connections between the diaphragm and the coupled masonry walls. In some cases, the 
diaphragm may yield, limiting the forces that can be transmitted to the in-plane walls. If interior 
partitions are connected to the partitions, the deformation, cracking damage and resulting energy 
dissipation can help limit the movement of the diaphragms. Such existing partitions or newly 
added partitions or moment frames are termed crosswalls. 

The ABK program identified two modes of behavior for in-plane loading on the piers in the 
unreinforced masonry walls: shear-critical behavior and rocking-critical behavior. Each of these 
modes of behavior could be found acceptable if demands were below the capacity. These two 
modes are included in the UCBC as well. FEMA 273 (FEMA, 1997a) and then FEMA 356 
(FEMA, 2000) expanded the characterization of in-plane behavioral modes into the more brittle 
force-controlled modes (toe crushing and diagonal tension) and the more ductile deformation

21-3 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

controlled modes (bed joint sliding and rocking). FEMA 306 and FEMA 307 identify a number 
of other in-plane modes and sequences of modes. 

The ABK program also established criteria for determining the acceptability of out-of-plane 
resistance of the unreinforced masonry walls. Important variables are the height-to-thickness 
(h/t) ratio, presence of crosswalls, and overburden (axial compression) pressure on the walls. In 
short, a stocky lower story wall with overburden pressure from floors and walls above that is 
driven by a diaphragm damped by crosswalls is less likely to buckle and fail out-of-plane. 

It is important to recognize that URM buildings vary substantially in structural layout and 
characteristics, and this can have a significant effect on seismic response. Fairly rectangular 
multistory residential, office, and commercial buildings often have relatively low story heights 
and many partitions that can serve as crosswalls. Walls adjacent to other buildings will usually 
be relatively solid. These buildings typically perform much better than structures like churches, 
which can have irregular plans, re-entrant corners, tall story heights, heavy walls, offset roofs, 
few partitions, and many windows. Churches can also be some of the most expensive structures 
to rehabilitate. 

21.3 Common Seismic Deficiencies and Applicable Rehabilitation 
Techniques 
URM buildings are generally considered to be one of the most hazardous building types. 
Significant property damage and loss of life have occurred in URM buildings during earthquakes 
around the world and in the United States. The primary deficiencies are due to unbraced 
parapets which can fall on adjacent pedestrian thoroughfares and poorly connected walls and 
diaphragms which can lead wall failure and loss of vertical support for diaphragms. See below 
for general discussion and Table 21.3-1 for a more detailed compilation of common seismic 
deficiencies and rehabilitation techniques for Building Type URM. 

Global Strength 

As shear wall buildings, global strength in URM buildings is dependent on the in-plane shear 
capacity of the walls. Relatively large seismic forces are needed to lead to life safety concerns 
with in-plane wall behavior, though cracking damage will occur in relatively moderate events. 
When walls are found to be deficient, new vertical lateral force-resisting elements can be added 
at interior locations or existing walls can be enhanced. At interior locations, new elements 
include wood structural panel shear walls, concrete shear walls, reinforced masonry shear walls, 
braced frames, and moment frames. At exterior locations, care must be taken to address relative 
rigidity concerns. Typically, concrete or shotcrete overlays are used to enhance the URM wall 
capacity. When the wall is highly punctured, braced frames or moment frames may be a viable 
option. The use of wood structural panel shear walls in buildings with masonry walls is 
permitted in new construction only in limited situations, such as one-story or two-story buildings 
with low story heights and no use of diaphragm rotation to resist loads, due to concerns about 
wood flexibility. Rehabilitation standards such as the 1997 UCBC and 2003 IEBC relax these 
restrictions significantly, though they do not permit the wood shear walls to resist lateral forces 
with other materials along the same line of resistance or when there are rigid diaphragms. Use of 
wood structural panel shear walls in rehabilitating masonry buildings should be carefully 
considered. 

21-4 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Table 21.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for URM Bearing Wall Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Global 
Strength 
Insufficient in-
plane wall 
strength 
Wood structural panel 
shear wall [6.4.2], [5.4.1] 
Concrete/masonry shear 
wall [21.4.8] 
Steel braced frame [7.4.1] 
Steel moment frame 
[21.4.9] 
Concrete wall overlay 
[21.4.5] 
Fiber composite wall 
overlay [21.4.6] 
Grouting 
Infill openings [21.4.7] 
Seismic 
isolation [24.3] 
Global 
Stiffness 
Configuration Soft story, weak 
story, excessive 
torsion 
Wood structural panel 
shear wall [6.4.2] 
Concrete/masonry shear 
wall [21.4.8] 
Steel braced frame [7.4.1] 
Steel moment frame 
[21.4.9] 
Load Path Inadequate or 
missing wall-todiaphragm 
tie 
Tension anchors 
[21.4.2] 
Shear anchors 
[21.4.2] 
Cross-ties and 
subdiaphragms 
[22.2.3] 
Supplemental 
vertical supports 
[21.4.11] 
Missing collector Add collector [7.4.2] 

21-5 


Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Table 21.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for URM Bearing Wall Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Component 
Detailing 
Wall inadequate 
for out-of-plane 
bending 
Exposed interfloor wall 
supports [21.4.3] 
Reinforced cores 
[21.4.4] 
Concrete wall overlay 
[21.4.5] 
Fiber composite wall 
overlay [21.4.6] 
Undesirable wall 
in-plane behavior 
mode 
Sawcutting to change 
shear mode to rocking 
mode 
Unbraced parapet Brace parapet [21.4.1] Remove parapet 
and improve 
roof-to-wall tie 
[21.4.1] 
Unbraced 
chimney 
Brace chimney [5.4.6] 
Infill chimney [5.4.6] 
Reduce 
chimney height 
[5.4.6] 
Remove 
chimney [5.4.6] 
Poorly anchored 
veneer or 
appendages 
Add ties [21.4.12] Remove veneer 
or appendages 
Diaphragms Inadequate 
in-plane strength 
and/or stiffness 
Add horizontal braced 
frame [22.2.9] 
Wood structural panel 
shear wall [6.4.2] 
Concrete/masonry shear 
wall [21.4.8] 
Steel braced frame [7.4.1] 
Steel moment frame 
[21.4.9] 
Wood structural panel or 
steel moment frame 
crosswall [21.4.10] 
Enhance existing 
diaphragm [22.2.1] 
Enhance woodframe 
crosswall [21.4.10] 

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Table 21.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for URM Bearing Wall Buildings 
Deficiency Rehabilitation Technique 
Category Deficiency Add New Elements 
Enhance Existing 
Elements 
Improve Connections 
Between Elements Reduce Demand 
Remove Selected 
Components 
Inadequate chord 
capacity 
Add steel strap or angle 
Diaphragms 
(continued) 
Excessive stresses 
at openings and 
irregularities 
Add wood or steel strap 
reinforcement 
Re-entrant corner Wood structural panel 
shear wall [6.4.2] 
Concrete/masonry shear 
wall [21.4.8] 
Steel braced frame [7.4.1] 
Steel moment frame 
[21.4.9] 
Collector [7.4.2] 
Foundation See Chapter 23 
[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 

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Global Stiffness 

URM bearing walls are generally quite rigid. When walls are solid or lightly punctured with 
window openings, global stiffness deficiencies are typically not an issue. In some buildings, 
though, facades facing the street can be highly punctured with relatively narrow piers between 
openings. In addition to lacking adequate strength, these wall lines may also be too flexible as 
well. 

Configuration 

Many commercial URM buildings will have a fairly open street faade at the ground level, 
leading to a weak and soft first story and torsional irregularities. This is usually addressed by the 
addition of a moment frame at the faade or another vertical lateral force-resisting element at 
some distance back from the faade. 

Load Path 

As noted above, it is the lack of adequate ties between the walls and diaphragms that is the single 
most significant deficiency in URM buildings. Rehabilitation measures include tension ties for 
out-of-plane forces and shear ties for in-plane forces. Bond beams are often employed for 
connections where the roof runs over the top of the walls. As a back-up vertical support system, 
supplemental vertical supports are added under trusses or girders where large gravity loads are 
concentrated on the wall in case the masonry is damaged locally. 

Component Detailing 

Since the masonry elements in Building Type URM are unreinforced by definition, they do not 
comply with modern ductile detailing requirements. Walls deemed susceptible to out-of-plane 
bending failures can be strengthened by strongbacks placed against them either on the outside or 
more commonly on the interior face. When preservation of finishes is critical, reinforced cores 
can be drilled and installed within the wall. Parapet bracing and chimney bracing are common. 
In some buildings, the exterior brick wythe will not be anchored back to the backing wall with 
mechanical ties or sufficient headers, and veneer ties are installed. 

Diaphragm Deficiencies 

Wood diaphragms may lack both strength and stiffness in URM buildings. This can be 
addressed by adding new interior elements to cut the diaphragm span or by enhancing the 
diaphragm itself with wood structural panel overlays. In some cases, such as sloped roofs, new 
horizontal braced frame diaphragms are added, in lieu of strengthening the existing diaphragm. 

Foundation Deficiencies 

Foundation deficiencies for existing elements are relatively uncommon in URM buildings. 
Foundation rehabilitation work usually is focused on the support for new lateral force-resisting 
elements that are added to the superstructure. 

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21.4 Detailed Description of Techniques Primarily Associated with 
This Building Type 
21.4.1 Brace or Remove URM Parapet 
Deficiency Addressed by Rehabilitation Technique 

Past earthquakes have consistently shown that unreinforced masonry chimneys and parapets are 
the first elements to fail in earthquakes due to inadequate bending strength and ductility. 
Parapets tend to have greater damage at midspan of diaphragms due to higher accelerations and 
displacements from the oscillating diaphragm. 

Description of the Rehabilitation Technique 

URM parapets can be braced or removed to minimize the falling hazard risk. Bracing is usually 
done with a steel angle brace. The brace is anchored near the top of the parapet and to the roof. 
The existing roof framing may need localized strengthening to take the reaction from the brace. 
Roof-to-wall tension anchors are typically part of parapet bracing. See Figure 21.4.1-1 for an 
example of parapet bracing. If the top of the parapet is removed, the vertical compressive stress 
on roof-to-wall anchors is reduced, so removing the parapet is often combined with adding a 
concrete cap or bond beam as part of the roof-to-wall anchorage. See Figure 21.4.1-2 for an 
example of parapet removal and addition of a concrete cap beam. See Section 21.4.2 for more 
details on wall-to-diaphragm anchorage. 

Design Considerations 

Research basis: No references directly addressing testing of parapet bracing have been 
identified. 

Parapet height: Codes such as the 2003 IEBC and 1997 UCBC provide maximum allowable 
height-to-thickness (h/t) ratios for parapets. The height is taken from the lower of the either the 
tension anchors or the roof sheathing. Requirements are more stringent in higher seismic zones. 
With the 1997 UCBC, for example, the h/t ratio in Seismic Zone 4 is 1.5, so for a typical 13 
thick, three-wythe wall, parapets taller than 19.5 above the roof-to-wall anchors require removal 
or bracing. In the 2003 IEBC, locations with SD1 > 0.4g have the same requirements. 

Fire protection: The original purpose of extending the masonry wall up to form a parapet was to 
help limit the spread of fire between the wood roofs of adjacent buildings. Removing a parapet 
must be coordinated with local building code fire safety requirements. 

Guardrails: Parapets often serve as a guardrail around a roof. Removing a parapet must be 
coordinated with local building code life safety requirements. 

Load path: When parapet bracing/roof-to-wall tension anchorage is the only rehabilitation 
technique, the out-of-plane load path can be incomplete, particularly when the roof joists are 
perpendicular to the brace. A more complete rehabilitation strategy includes developing the 
parapet and tension anchorage forces back into the roof diaphragm through the use of 
subdiaphragms and cross-ties. 

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Techniques for the Seismic Rehabilitation of 
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Figure 21.4.1-1: Parapet Bracing 

Detailing and Construction Considerations 

Parapet anchorage types: Drilled dowels connecting the top of the bracing to the masonry can 
be with through bolts or adhesive anchors. See Section 21.4.2 for detailed discussion of drilled 
dowels. 

Top angle: Figure 21.4.1-1 shows a continuous angle running between braces in the roof. This 
angle can be used to span between braces to reduce the number of bracing points. It also 
increases redundancy over a localized connection of the brace to the parapet. 

Load in the roof framing: The vertical reactions at the base of the brace are typically resisted by 
roof framing. In Figure 21.4.1-1 the added blocking beneath the base of brace workpoint helps 
to engage three joists in resisting vertical loads. Tall parapets can generate substantial brace 
forces that existing wood roof joists may not be able to resist. Additional joists can be added, or 
more braces can be used to distribute the load. Horizontal loads from the brace are distributed by 
the blocking and new wood structural panel. 

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Techniques for the Seismic Rehabilitation of 
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Figure 21.4.1-2: Parapet Removal and Concrete Cap Beam 

Waterproofing at the roof: The brace anchor at the roof needs to attach to the structural framing 
members, so a penetration through the roof membrane will occur that needs proper 
waterproofing design by a qualified contractor or design professional. Often, parapet bracing 

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Techniques for the Seismic Rehabilitation of 
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and roof-to-wall ties and even roof diaphragm sheathing rehabilitation activities are combined 
with roofing replacement given the cost effectiveness of combining the work. 

Cost and Disruption Considerations 

Adding parapet bracing and roof-to-wall tension anchors provide some of the most effective 
seismic rehabilitation for reducing life safety risks. As a result, some communitiessuch as San 
Franciscopassed parapet safety ordinances requiring mandatory mitigation many years ago. 
Disruption is typically relatively low since occupants can remain in place. Combining parapet 
bracing and roof-to-wall ties and even roof diaphragm sheathing rehabilitation activities with 
roofing replacement can significantly reduce the total cost of the work. Disruption can increase 
noticeably if the roof has to be removed for installation. 

Proprietary Issues 

There are no proprietary concerns with parapet bracing, other than use of proprietary anchors as 
part of the assemblage. See Section 21.4.2. 

21.4.2 Add Wall-to-Diaphragm Ties 
Deficiency Addressed by Rehabilitation Technique 

Inadequate or missing shear and tension connections between the unreinforced masonry bearing 
wall and the wood floor or roof. 

Description of the Rehabilitation Technique 

The most significant deficiency in URM bearing wall buildings is the lack of an adequate 
positive (i.e. mechanical) tie between the masonry walls and the floor and roof diaphragms. Ties 
are usually separated into two categories: tension ties and shear ties. Tension ties transfer out-ofplane 
inertial loads perpendicular to the face of the masonry back into the diaphragm. Shear ties 
transfer loads from the diaphragm into the wall where they are resisted by in-plane action of the 
wall. Tension ties help keep the walls from falling away from the diaphragms; shear ties help 
keep the diaphragm from sliding along parallel to the wall. Ties are assemblages that consist of 
both the anchorage to the wall (shown in detail in Figures 21.4.2-1 and 21.4.2-2) and the 
anchorage back into the diaphragm (shown in the subsequent figures). 

Design Considerations 

Research basis: The focus of wall-to-diaphragm testing to date has been on the anchorage to the 
masonry and has been done primarily by manufacturers. Paquette, Bruneau and Brzev (2003) 
tested a specimen of a small full-scale one-story building with roof-to-wall ties, but the focus of 
the work was on wall and diaphragm response. 

Anchor types and capacities: The 1997 UCBC and 2003 IEBC provide prescriptive values for 
tension and shear bolts meeting certain requirements. These are for a 2-1/2 diameter hole filled 
with nonshrink grout approach that is typically no longer used. The ICBO and now ICC 
evaluation report process has standardized procedures for vendors supplying adhesive ties for use 
in brick masonry. Three installations are included in most vendors ICC Evaluation Service 
reports, and they have standardized installation techniques and capacities. Adapted versions of 
these installations are shown in Figures 21.4.2-1 and 21.4.2-2. Figure 21.4.2-1A shows a 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Figure 21.4.2-1: Drilled Dowels 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Figure 21.4.2-2: Through Bolt Anchor 

combination drilled dowel that can be used for resisting both tension and shear forces. It is 
drilled into the wall at a 22.5 degree angle from horizontal at least 13 into the wall. The angle 
allows the dowel to engage more courses of brick, theoretically improving the reliability. At the 
allowable stress design (ASD) force level, it is good for 1200 lbs in tension and 1000 lbs in 
shear. Figure 21.4.2-1B shows a drilled dowel used only for resisting shear forces. It goes in 8 
deep into the masonry and is good for 1000 lbs at the ASD level. Figure 21.4.2-2 shows a 
special through bolt anchor using a steel sleeve in the first 8 that can take tension and shear and 
has the same values as the combination anchor. These ICC capacities are typically used in 
design; they come with a number of restrictions and requirements such as quality of masonry. 
When higher values are needed, proof testing can be undertaken. In the ICC standards for both 
shear and tension testing (ICC-ES, 2005) of adhesive anchors that manufacturers must use to 
obtain ICC qualification, allowable stress design capacities are the lower of prescriptive values 
and the average ultimate test value divided by a safety factors of 5. 

It should be appreciated that the prescriptive values in the UCBC, the IEBC, and ICC Evaluation 
Service reports are based on tests of the drilled dowel itself, not the full elements of the detail. 

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Techniques for the Seismic Rehabilitation of 
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Capacities for nails, wood structural panels, bolts in wood and straps come from typical code 
provisions. 

Detailing and Construction Considerations 

There are many issues to consider in detailing for tension and shear ties. These include the 
following: 

Aesthetics: Anchors that go all the way through the wall have a visible bearing plate on the 
exterior face, such as shown in Figure 21.4.2-2. There are simple circular or octagonal plates 
that can be purchased or fabricated. Some manufacturers make plates with a countersunk hole 
and use flathead bolt heads to reduce the surface projection. When the exterior face is stucco, a 
plate with a countersunk hole can be recessed into the stucco or just into the masonry and 
refinished with stucco so it is hidden. Special cast anchors can be made if there is a desire to 
match an historic exposed cast iron anchor. When the anchor plate approach cannot be used, 
drilled dowels are used such as those shown in Figure 21.4.2-1. 

Nonshrink grout vs. chemical adhesive: Early ties used cementitious nonshrink grout. They 
required larger diameter holes (such as 2-1/2) to be cored in the masonry to place the grout. A 
number of vendors have now created special chemical adhesives and tools that have optimized 
the process. Standard details use  diameter threaded rods in 1 diameter holes, though other 
sizes can be used, depending on manufacturer requirements. The typical installation approach is 
to drill the hole; clean it with a brush and compressed air; fill a nylon, carbon, or stainless steel 
screen tube (which looks like a test tube made out of wire mesh) with adhesive; place the screen 
tube into the hole; and then push the rod into the screen tube forcing the adhesive out of the tube 
into the annulus between the tube and the masonry. Figures 21.4.2-1 and 21.4.2-2 show the 
anchorage using chemical adhesives and screen tubes. 

Chemical adhesive types: There are many different types of chemical adhesives, though most 
are epoxy. Epoxy products have the longest track record. Some vendors have begun to produce 
other types of chemicals. Key issues when considering an adhesive are the length of time the 
adhesive has been in use, the extent and quality of the testing, the ability to bond to damp or 
water filled surfaces, setting time, cost, the heat deflection temperature (an ASTM test method 
for quantifying the loss of strength as ambient temperature rises), and the capacities shown by 
test results. Most modern adhesives use two-component pre-packaged assemblies, rather than 
bulk products used in the past. This reduces the risk of improper mixing and not developing the 
adhesive to its proper strength. When adhesives are curing, the off-gassing can be unpleasant, 
and proper ventilation procedures are necessary. 

Dowel material type: Threaded rod is commonly specified as ATSM A36 all-thread rod. It is a 
relatively ductile material, with a minimum yield strength of 36 ksi and ultimate strength of 60 
ksi. When higher strength material is needed (which is rare), ASTM A193, Grade B7 threaded 
rod can be used with a minimum yield strength of 105 ksi and ultimate strength of 125 ksi. 
Rebar can be used as well, but this is not typically done in ties that connect to the wood 
diaphragms since the threaded connection is needed. Threaded rod is sometimes supplied with 
oil on it. This must be solvent cleaned, so that proper bonding with adhesives can occur. 

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Techniques for the Seismic Rehabilitation of 
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Access: Installation of ties can be done either from below the diaphragm or above. Figure 
21.4.2-3 shows installation of floor-to-wall tension ties from below. Figure 21.4.2-4 shows 
installation occurring from above the floor. Figure 21.4.2-5 shows installation of floor-to-wall 
shear dowels from above. Similar details are contained in Rutherford & Chekene (1990), 
SEAOSC (1982) and SEAOSC (1986). The choice of whether to install from above or below 
depends on whether there are finishes that need to be avoided, whether diaphragm strengthening 
is being done, and what type of diaphragm strengthening is planned. If there is a special plaster 
ceiling to be avoided, then access and installation would proceed from above. If there is no 
plaster ceiling and the floor or roof diaphragm is not being modified or is being enhanced by 
adding a wood structural panel overlay from above, then access and installation for wall-
diaphragm ties would be from below. Angled dowels (see section below) installed from below 
can be angled upwards rather than the typical downward angle, provided non-sag adhesives are 
used. 

Joist direction: Framing in most buildings is orthogonal so that joists or rafters are either 
perpendicular or parallel to the in-plane direction of the wall. Installations where the joists are 
perpendicular to the wall are easier to make; installations where the joists are parallel involve 
blocking and more complicated details. Figures 21.4.2-3 to 21.4.2-5 show variations for joist 
orientation. 

Special issues at the top of the wall: In most URM buildings, the wall continues up past the roof 
forming a parapet that provides fire protection and serves as a guardrail during roof maintenance, 
as described in Section 24.4.1. In some buildings, though, the roof continues over the top of the 
wall. In these situations, the roof might be relatively flat or sloped. As a result, special issues 
arise. First, there is reduced overburden pressure at the top of the wall, reducing the reliability of 
drilled dowels. Second, eccentricities become more significant, such as the vertical eccentricity 
between the roof diaphragm and the top of the masonry. Making reliable connections between 
walls in these situations can be particularly challenging and is usually dependent on the specific 
geometry and characteristics of the existing details. A common strategy is to employ a concrete 
bond beam at the top of the wall. This ties the wall together, serves as a collector and chord, 
increases redundancy and often simplifies details. Figure 21.4.2-6 shows a bond beam placed on 
top of an existing wall under the roof framing. This is possible when the wall is wide, and there 
is sufficient distance between the masonry and rafter. Figure 21.4.2-7 shows an alternative when 
there is insufficient clearance between the rafter and top of wall that involves removing the top 
two courses of masonry to gain room for the bond beam. 

Eccentricity: It is desirable to minimize the eccentricities in a connection. Figure 21.4.2-8 
illustrates the issue and some alternate approaches with floor-to-wall tension ties. Figure 21.4.28A 
shows a common tension tie detail in plan view where a tie-down anchor is connected to the 
side of an existing joist. The plan offset between the drilled dowel at the center of the tie-down 
where load is applied and the center of the joist where it is resisted times the force is a moment 
that must be resisted by the joist in weak way bending. This stress can be quite significant. 
Figure 21.4.2-8B shows an alternative where two tie-downs are used to make a connection that is 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Figure 21.4.2-3: Tension Anchors Installed from Below the Floor 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Figure 21.4.2-4: Tension Anchors Installed from Above the Floor 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Figure 21.4.2-5: Floor-to-Wall Shear Anchors 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Figure 21.4.2-6: Bond Beam at a Sloping Roof 

Figure 21.4.2-7: Bond Beam at a Sloping Roof with Limited Clearance 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Figure 21.4.2-8: Tension Tie Connection Issues 

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Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

more concentric. This detail, however, puts a large number of screws into the existing joist, so a 
sistered joist is shown. Adding the sister also permits the nailing into the diaphragm to be into 
each joist, reducing the nailing demand on the joists. Bolted tie-downs, instead of tie-downs 
with screws, can be used with through bolts placed in double shear. Traditional bolted tie-downs 
have greater slip than the more recent tie-downs using screws. There are proprietary connectors 
using tubes as tie-downs on each side without oversize holes that bolt eliminate eccentricity and 
reduce bolt slip. Both Figure 21.4.2-8A and 21.4.2-8B have dowels adjacent to the joist. This 
means the dowel will enter the wall next to or in the weakened area of joist pocket and at the end 
of new blocking used for shear transfer, where there is insufficient end distance to use the dowel 
as a shear tie. Figure 21.4.2-8C shows a V-strap detail where the drilled dowel is placed between 
joists, away from the joist pocket and with plenty of end distance. When the strap is in tension, 
forces perpendicular to the joists are produced that are resisted by the added blocking and plate 
washers. 

Truss anchorage: In some URM buildings, there will be large gravity elements that bear on the 
wall, such as girders or trusses. These also become concentrated points of stiffness in the 
diaphragm. Since the relative rigidity of the elements cannot be easily quantified, it is usually 
prudent to use an enveloping or belt and suspenders approach of assigning demand, so that 
typical anchors between trusses take the uniform load and the ties connecting the wall and 
trusses take additional load. 

New ties vs. reuse of existing ties: In many older URM buildings, there are existing ties called 
government or dog anchors. These anchors typically only occur in the direction where the 
joists are perpendicular to the face of the wall, and they may not be at sufficient spacing. The 
1997 UCBC and 2003 IEBC permit use of these anchors as wall-to-diaphragm tension anchors if 
tested in accordance with certain standards and capacities are sufficient. 

Dowel spacing and edge distance: The 1997 UCBC and 2003 IEBC have maximum spacing 
requirements on shear and tension dowels. When walls become thick, the out-of-plane demands 
and the relatively low ICC Evaluation Service report capacity values can lead to fairly tight 
spacing of dowels. The UCBC and IEBC do not have minimum spacing requirements. From a 
practical point of view, dowels should not be placed closer than 12 o.c. Some ICC reports 
provide minimum spacing limits as well, like those commonly employed for drilled dowels in 
concrete. For one vendor, these spacing limits are 16 o.c. in the horizontal and vertical 
direction, and there is 16minimum for edge distance as well. 

Corrosion considerations: Drilled dowels are typically installed from the interior. The masonry 
cover and epoxy serve as corrosion protection, so mild steel anchors are typically considered 
sufficient. For increased corrosion protection, stainless steel dowels and screen tubes can be 
used. When through bolted connections are installed, there is a more direct path for moisture 
intrusion. The anchor plate can be painted with exterior grade paint, galvanized or be made from 
stainless steel, and the through bolt can be made from stainless steel as well. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Screen tubes: The purpose of the screen tube is to prevent loss of epoxy into cracks or unfilled 
collar joint voids within the wall. Screen tubes vary somewhat from vendor to vendor and 
should be considered part of the manufacturers assembly. Nylon screen tubes have begun to be 
supplied by many vendors as they are more economical than stainless steel and more corrosion 
resistant than carbon steel. They do have a much larger coefficient of thermal expansion than 
both steel screen tubes and masonry. 

Hollow masonry: Anchorage of hollow clay tile, ungrouted concrete masonry units and other 
hollow masonry systems to diaphragms is particularly challenging. When forces are large, 
grouting in the region of the anchor is usually required. When forces are small, use of screen 
tubes may be acceptable. The screen tube is filled with adhesive, inserted into the wall and as 
the dowel is pushed into it, the adhesive seeps through the screen tube forming a key behind the 
face shell of the masonry. Capacities are small and the connection is nonductile. This type of 
connection may be viable for out-of-plane wall strengthening (see Section 21.4.3) where the 
demands are lower, but it is not recommended for wall-to-diaphragm connections. Figure 
21.4.2-9 shows a method of connecting a floor to an ungrouted CMU wall. Even in ungrouted 
CMU, a grouted bond beam is usually found beneath the floor, and it helps provide bearing 
support for the floor joists. Figure 21.4.2-9 involves locally grouting the courses at and just 
above the floor to install a new anchor. Figure 21.4.2-10 shows an alternative that avoids 
working from above and uses the existing bond beam. Sistering and a nailer help get the new 
anchor to the proper elevation. If a grouted bond beam is not present, it may be necessary to 
create one to make the proper anchorage, similar to the top courses in Figure 21.4.2-9. 

Drilling: Holes need to be drilled with a rotary drill or a rotohammer drill with the percussion 
setting turned off to limit vibration into the wall. This can slow drilling significantly. In some 
cases, coring with a diamond tipped blade is more efficient. This may be the only way some 
hard masonry, like granite, can be drilled. Sometimes water is used to cool the bit, and the slurry 
produced by the water, mortar and masonry can stain the face of the wall. 

Cost/Disruption 

Considerations for cost depend on the number, type and depth of dowels; the difficulty of access; 
and the extent of finishes that are impacted. Through bolts are usually less expensive than 
adhesive anchors. 

Drilling is loud and can be disruptive to occupants. Typically, either the floor or ceiling has to 
be removed to install the dowels. Thus, it is usually not practical to install dowels in occupied 
rooms, though the work can be phased by building area so disruption is minimized. 

Proprietary Issues 

Values for anchor capacity come from individual vendors, but there are no known concerns with 
use of a properly procured product. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Figure 21.4.2-9: Wall-to-Floor Tension Tie in Hollow Masonry 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Figure 21.4.2-10: Wall-to-Floor Tension Tie in Hollow Masonry Alternate 

21.4.3 Add Out-of-Plane Bracing for URM Walls 
Deficiency Addressed by Rehabilitation Technique 

Inadequate out-of-plane bending resistance of an unreinforced masonry wall. 

Description of the Rehabilitation Technique 

Two types of bracing can be used: diagonal braces that reduce the effective height of the 
masonry wall (Figure 21.4.3-1A) and vertical braces or strongbacks that span the full height of 
the inside face of the wall (Figure 21.4.3-1B). Vertical braces can be surface mounted or, when 
aesthetic considerations are paramount, recessed into the wall; see Figure 21.4.3-2. 

Design Considerations 

Research basis: The most comprehensive set of testing done to date on out-of-plane response of 
URM walls was part of the ABK research program in the 1980s, and it is documented in ABK 
(1981c). Full-scale, dynamic testing of 20 wall specimens was conducted. Specimens were 6 
wide, 10 to 16 tall, and had height-to-thickness (h/t) ratios that varied from 14 to 25. 
Superimposed axial loads were varied; and materials included brick, grouted CMU, and 
ungrouted CMU. 

H/t limits: It is tall, narrow walls that have been found to be susceptible to out-of-plane wall 
demands. The 1997 UCBC and 2003 IEBC provide maximum h/t requirements. Walls with 
larger h/t ratios must be braced. 

Spacing: For strongbacks, such as shown in Figures 21.4.3-1B and 21.4.3-2A, the maximum 
spacing requirements are set by the 1997 UCBC or 2003 IEBC at the minimum of 10 feet or half 
the unsupported height of the wall. For diagonal braces, the maximum spacing is set at 6 feet. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Figure 21.4.3-1: Exposed Out-of-Plane Wall Bracing 

Stiffness: For strongbacks, such as shown in Figures 21.4.3-1B and 21.4.3-2A, the 1997 UCBC 
limits deflection of the wall at ASD demands to one tenth of the wall thickness. This is not a 
particularly stringent requirement. Say that the first story of a multistory building in Seismic 
Zone 4 is 13 thick and 18 tall and its resulting h/t ratio of 16.7 exceeds the h/t limit of 16 in the 
UCBC. Bracing would be need to be stiff enough to keep deflections down to 10% of 13 or 
1.3. This is L/166, which is comparatively low to most masonry design requirements, which 
are typically L/360 or higher, up to even L/600. Kariotis (1982) notes that the goal of a flexible 
vertical brace is to keep the brace elastic and provide a predictable restoring force during cracked 
excursions of the masonry wall. For diagonal braces, the UCBC encourages detailing to 
minimize vertical deflections. 

Diagonal braces loading vs. bracing the wall: If the roof deflects downward on a diagonal brace, 
a horizontal reaction is imparted to the wall. One concern with diagonal braces is that vertical 
vibration of the roof in an earthquake can contribute to the out-of-plane inertial forces on the 
wall. This concern, combined with the difficulty of making the roof stiff enough for against 
vertical deflections, makes vertical bracing a preferred engineering choice over diagonal bracing. 

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Techniques for the Seismic Rehabilitation of 
Existing Buildings: FEMA 547 Chapter 21  Type URM: Unreinforced Masonry Bearing Walls 

Figure 21.4.3-2: Vertical Bracing Alternatives 

Recessed steel and concrete and surface-mounted concrete: Provisions in the 1997 UCBC and 
2003 IEBC do not explicitly consider the approaches shown in Figures 21.4.3-2B, 21.4.3-2C and 
21.4.3-2D. These approaches are unusual, but they can be used when a more sensitive aesthetic 
approach or higher loads are needed. 

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Detailing and Construction Considerations 

Materials: Braces are typically done with steel as shown in Figures 21.4.3-1 and 21.4.3-2A, but 
strongbacks can also be done with wood posts or with concrete pilasters (Figure 21.4.3-2C). 

Aesthetics: Figure 21.4.3-1 shows exposed braces. This is the least expensive approach and is 
appropriate for certain occupancies. When there is architectural desire to hide the steel, the 
bracing can be furred at added cost and impact on the usable space. To minimize the impact on 
the space, the vertical brace can recessed into a cavity cut in the wall with either a steel or a 
concrete member. See Figure 21.4.3-2. Recessing the steel or concrete requires significantly 
more work and raises the potential for cracking to propagate from the inside of the recess to the 
masonry face. 

Strongback anchor spacing: Figure 21.4.3-1B shows only a central anchor at midheight of the 
wall. Often demand/capacity ratios for anchorage to the wall with through bolts or drilled 
dowels (see Section 21.4.2) will dictate a tighter spacing of anchors. 

Floor/roof framing capacity: Figure 21.4.3-1 shows anchorage to joists oriented perpendicular 
to the wall. When joists are parallel to the wall, the horizontal anchorage force must be 
developed out into the diaphragm. In Figure 21.4.3-1A, the existing roof beams may need to be 
strengthened to provide adequate strength to resist downward loading. 

Hollow masonry: Figures 21.4.3-1 and 21.4.3-2 apply to solid masonry. When the existing 
masonry is hollow, alternative connection methods are needed. Figure 21.4.3-3 shows use of 
vertical concrete ribs. A chase is created by removing the face shell on one side of the wall. 
Reinforcing steel is added and then grout or concrete fill. There is typically insufficient space 
for ties. This approach is messy and noisy. Figure 21.4.3-4 shows an alternative where steel 
strongbacks are bolted to the wall with either drilled dowels or through bolts. The screen tube 
anchor of Figure 21.4.3-4A relies on mechanical keying action from the spreading adhesive to 
engage the face shell. The capacity is limited to the face shell of the masonry and can be quite 
low, in the low hundreds of pounds at allowable stress design levels. It is also nonductile as the 
failure mechanism is spalling of the face shell. The through bolt in Figure 21.4.3-4B provides 
increased capacity and locally grouting in the anchor provides additional capacity. 

Cost/Disruption 

Diagonal bracing is usually less expensive, but is considered less reliable than vertical bracing. 
Furring can be used to cover the braces at added cost. Exposed braces are typically less 
expensive than more architecturally sensitive alternatives like recessed vertical braces or 
reinforced cores (See Section 21.4.4). Installation of bracing is fairly disruptive since it must 
occur around the entire perimeter; and it involves drilled dowels, and accessing and connecting 
to horizontal diaphragms. 

Proprietary Issues 

There are no known proprietary concerns with bracing of URM walls. 

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Figure 21.4.3-3: Concrete Ribs in Hollow Masonry 

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Figure 21.4.3-4: Connection of Strongback to Hollow Masonry 

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21.4.4 Add Reinforced Cores to URM Walls 
Deficiencies Addressed by Rehabilitation Technique 

Inadequate out-of-plane capacity and in-plane capacity of unreinforced masonry wall. 

Description of the Rehabilitation Technique 

Installing the reinforced core involves drilling a core from the roof down the inside of an 
unreinforced masonry wall. A steel reinforcing bar and grout are placed inside the hole to 
increase the wall strength. See Figures 21.4.4-1A and 21.4.4-1B. This process is used to avoid 
the aesthetic impact of exposed bracing described in Section 21.4.3. 

Design Considerations 

Research basis: The original research at CSU Long Beach and North Carolina State University 
for reinforced cores is summarized in (Plecnik, Cousins, and OConner, 1986) and Plecnik 
(1988). It covered both out-of-plane and in-plane loading. Subsequent vendor tests for in-plane 
loading were done at UC Irvine but have not been published. More recent in-plane testing is 
summarized in Abrams and Lynch (2001). 

Out-of-plane capacity: When reinforced cores are used for enhancing out-of-plane bending 
capacity, the wall is analyzed as a reinforced masonry element. Some engineers have used a 
traditional allowable stress design code format for masonry. Another common approach is to use 
factored design methods. Plecnik, Cousins and OConner (1986) provided an ultimate strength 
design formulation. As with concrete design or typical reinforced masonry design, the 
compressive strength of the masonry, fm, is needed. Default values are available in FEMA 356, 
but it is important not to over-reinforce the masonry section and cause a brittle failure of the 
masonry, so it is often prudent to obtain the masonry strength. 

In-plane capacity: Plecnik, Cousins and OConner (1986) tests showed significant increase in in-
plane loading from the reinforced cores, but a design methodology was not provided. Breiholtz 
(1987) suggested using the test results as well, but did not provide a complete design 
methodology. One design approach is simply to extrapolate the test results on a per lineal foot 
basis. Another is to consider the vertical bars as the tie element in a strut-and-tie methodology, 
with diagonal struts in the masonry connecting the cores. 

Post-tensioned masonry: Post-tensioned masonry has been used in a few instances. The goal of 
post-tensioning the bars can be to add compressive stress to the masonry wall to increase the 
effective shear stress, since unreinforced masonry shear capacity formulations, such as those in 
FEMA 356 or the ICBC, provided increased shear strength with higher compressive strength. 
Recent research by Rosenboom and Kowalsky (2004) provides some results from cyclic testing, 
but the specimens have central cavities filled with grout rather than solid multi-wythe brick, and 
design equations are not provided. 

Detailing and Construction Considerations 

Detailing and construction considerations for reinforced cores include the following. 

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Figure 21.4.4-1A: Reinforced Cores 

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Figure 21.4.4-1B: Plan Detail of Reinforced Core in Masonry Wall 

Wet vs. dry drilling: Traditionally, holes cored in masonry were done similarly to those in 
concrete, using diamond tipped coring bits cooled by water. The slurry created by the water and 
brick dust can lead to staining of sensitive surfaces. Reinforced cores gained popularity when 
drilling companies developed specialized drilling equipment that did not need water to cool the 
bit. In many cases, this involves coring drills that rotate quite slowly compared to traditional 
coring equipment. The material within the core comes out in cylindrical chunks or in small 
pieces of debris or dust that are vacuumed into debris containers. 

Reducing leaks: To minimize leaks during wet drilling and during grouting, loose masonry 
should be repaired and cracks repointed. To limit the extent of repointing, consideration can be 
given to monitoring the location of dust clouds that escape cracks during dry drilling and 
repointing those locations. 

Obstructions: Drilling progresses most rapidly if the masonry is neither too hard nor too soft and 
is relatively homogeneous. Encountering wood debris inside walls or metal veneer ties can slow 
or stop the drilling. 

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Angled drilling: Drilling is typically done from the top and straight down. Occasionally, special 
situations arise where angled drilling might be necessary. This can be done, but requires much 
greater skill from the driller. 

Hole diameter: Hole diameters are typically about four inches, but can range from three to six 
inches. 

Drilling tolerances: The wider and shorter the wall, the easier it is to drill because it provides 
better tolerance against drilling inaccuracy. Tolerances of about 2 in reasonably tall walls are 
usually achievable. A simple way to check this is as follows. Say the hole is four inches in 
diameter. Attach a small penlight flashlight to the end of a string, creating a lighted plumb bob. 
Drop the plumb bob down from the center of the hole at the top. If it does not hit the side of the 
cored hole at any point on the way down, a two-inch tolerance has been met. 

Bar material type: The reinforcing bars used in reinforced cores are typically regular ASTM 
A615 mild steel, as they are protected by the grout. For increased corrosion protection, stainless 
steel or epoxy coated rebar can be used. If post-tensioned center coring is done, high strength 
ASTM A722 threaded bars can be used. 

Bar size: Bar size depends on the demand/capacity ratios, but typically ranges from #5 to #8. 

Centralizers: In order to keep the rebar centered in the hole, a plastic centralizing wheel is used 
by some engineers. Others consider this an obstruction limiting the flow of grout. 

Grout type: The original research by Plecnik, Cousins and OConner (1986) evaluated several 
grouts and concluded that a formulation using polyester grout provided the best dispersion into 
the masonry. Polyester has some offgassing concerns from styrene vapors and requires the hole 
to be dry before installing the grout. It also does not have the long-term track record of other 
more widely used grout materials. Some engineers, as a result, use a high quality nonshrink 
cementitious grout. With cementitious grouts, the hole needs to be prewetted prior to grout 
placement. 

Grouting process: Center coring can be done in multistory buildings, so the depth of holes can 
get quite large. A tremie grouting technique can be used to assure placement of grout. A grout 
tube can be tied loosely to the bar and/or centralizer(s) as the bar is lowered into the hole. As 
grout is pumped into the hole, the tube is slowly withdrawn as the level of grout rises. 

Verification ports: Grout will leak into voids in the masonry. To confirm that the grout is rising 
in the core, horizontal holes can be drilled into the wall. When the grout reaches the port, the 
port is plugged, and the grout is allowed to continue to rise to the top of the core. 

Bottom of holes: Coring usually goes into the foundation. In a concrete foundation, there will 
not be any place for water used to cool the bit in wet coring to escape. The hole can be 
vacuumed out, or the core can be continued all the way down to the bottom of the foundation. 

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Post-tensioning: When reinforced cores are post-tensioned, several additional issues come into 
play. First, grouting is done in two stages. The first stage is in the foundation where the post-
tensioning is anchored. After the grout cures and the bar is stressed, the second stage of grouting 
occurs up to the top of the hole. Post-tensioning vendors provide proprietary anchorage 
hardware for the bar at the top of the wall. A concrete cap or bond beam may be necessary or 
desirable to distribute the load on the top of the wall to reduce the stress on the masonry. 

Access to top of wall: The top of the wall must be accessible to drilling equipment. Temporary 
scaffolding or work platforms will usually need to be erected adjacent to the hole. Bracing of 
drilling equipment back to the wall and a point in the roof is necessary to keep the drill plumb. 

Spacing: Reinforced core spacing will depend on demand/capacity ratios, but a minimum 
spacing of six to ten feet is desirable. 

Cost/Disruption 

Adding reinforced cores can be considerably more expensive than exposed bracing, so it is 
usually only performed in historically and architecturally sensitive buildings. Since the core is 
placed inside the wall, the disruption to interior and exterior faces is limited to sealing cracks, 
access to the roof to place the drilling equipment, and drilling noise and vibration. 

Proprietary Issues 

Research for reinforced cores is in the public domain. Some drilling contractors reportedly have 
patents on certain types of proprietary drills. Some of the terms used with the process have been 
trademarked by some of the first engineers to implement the technique. As a result, the generic 
term reinforced core is used in this document. 

21.4.5 Add Concrete Overlay to Masonry Wall 
Deficiency Addressed by Rehabilitation Technique 

Improving inadequate in-plane wall capacity is the primary purpose of a new concrete overlay, 
but the concrete can also improve inadequate out-of-plane bending capacity. 

Description of the Rehabilitation Technique 

New concrete is applied against an existing unreinforced masonry wall to increase the shear 
capacity of the wall. The new concrete is attached to the old wall with adhesive anchors and can 
either be cast-in-place concrete or sprayed-in-place. In rehabilitation work, sprayed concrete, 
known as shotcrete, is more commonly employed than cast-in-place construction, since the 
existing wall provides the back-side form. The thickness of the new concrete varies with 
strength requirements, but it is usually from four to 12 inches. See Figure 21.4.5-1. 

Design Considerations 

Research basis: A fair amount of research in the use of shotcrete overlays on masonry has been 
done. A summary is given in El Gawady, Lestuzzi and Badoux (2004). Early diagonal tension 
testing was done by Kahn (1984), and more recently static cyclic tests were done by Abrams and 
Lynch (2001). Kahn (1984) showed significant increases in strength from the shotcrete and that 
adding drilled dowels between the shotcrete and masonry or an epoxy bonding agent did not lead 

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Techniques for the Seismic Rehabilitation of 
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Figure 21.4.5-1: Concrete or Shotcrete Wall Overlay 

to significant improvements. A saturated masonry surface was recommended. Testing by 
Abrams and Lynch (2001) aimed at increasing the shear capacity to lead to flexural yielding of 
the tension bars in the shotcrete. Strength increased by about a factor of 3, but displacement 
capacity did not increase. 

Design criteria: When a concrete overlay is used, there are several common force-based design 
approaches for the wall, due to the relatively high strength of the concrete compared to the 
masonry. One is to take 100% of the demand tributary to the strengthened wall line in the 
concrete overlay itself and ignore the masonry. While this may sound conservative, it can mean 
that the masonry will be significantly damaged before the concrete ever sees the majority of its 
design load. Another approach is to share the load, by relative rigidity, between the masonry and 
the concrete. When this is done, both the masonry and the concrete must be checked to confirm 
they are not overstressed. The most conservative approach is to use the overlay to resist 100% of 
the tributary load, but to also check that the masonry can resist the loads it will actually attract. 
Displacement-based design approaches inherently consider the relative rigidity of the concrete 
and masonry, but they are less commonly employed. 

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Discretely applied overlays: The URM walls to which the overlay is applied are typically 
punctured with window and door openings. It can be tempting to apply the concrete to wide 
piers. The comparatively high strength of the concrete means it can take high loads, but it is 
unlikely to have sufficient stiffness to actually attract the load it was intended to take. 
Eventually, though, when the masonry cracks, the load will find the concrete, but this can lead to 
significant cracking at the ends of the concrete in the masonry spandrels. See Figure 21.4.5-2A. 
This can be addressed by spreading out the influence of the overlay by using top and/or bottom 
spandrels or grade beams, such as shown in Figure 21.4.5-2B. Alternatively, drilled piers can be 
placed at the ends of the new walls to add stiffness, as in Figure 21.4.5-2C. Or most simply, a 
continuous overlay can be used, enabling a reduced thickness and consistent finish surface, as 
shown in Figure 21.4.5-2C. 

Collector load transfer pathways: With discretely applied overlays in the field of the masonry 
wall, the question arises of how the load in the floors and the masonry wall will reach the 
overlay. Some engineers ignore this issue and assume the masonry wall will serve as the 
collector. Others provide an explicit steel collector in the edge of the diaphragm or a bond beam 
on top of the wall. 

Out-of-plane load resistance: When the overlay is added to the wall, its additional inertial load 
must be considered in the seismic weight of the structure. Out-of-plane anchorage requirements 
at the overlays are thus larger, and wall-to-diaphragm tie spacing often decreases there. 

Detailing and Construction Considerations 

Detailing and construction considerations for concrete overlays include the following. 

Drilled dowel spacing: It is common practice to connect the overlay to the masonry wall with 
drilled dowels. The drilled dowels transfer shear between the two materials, and they also serve 
as out-of-plane ties for the masonry. Spacing of two feet to three feet on center is typical. Good 
detailing involves drawing an elevation and showing the location of dowels around openings 
since the nominal spacing will typically change there. 

New foundation at the base of wall: The base of the shotcrete can be set on the ledge of the 
existing footing if conditions permit or a new footing can be provided. The added load from the 
shotcrete and distribution of stresses on the existing footing must be considered. See Section 

23.6.2 for issues involved in adding a new footing next to an existing footing. 
Interface between the wall and diaphragms: Where the new wall meets the existing floors is 
usually the location where special consideration must be given to detailing. Conditions where 
the floor joists are parallel to the wall are the easiest to address. The first and/or second joists in 
from the masonry are removed to install the overlay, a ledger is placed back, and the floor 
sheathing run up to the ledger. If shotcrete is used, sufficient clearance must be provided to 
avoid shadowing during spraying. This should be checked as part of the preconstruction test 
panel. 

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Figure 21.4.5-2A: Potential Damage at Ends of Narrow Concrete Overlay 

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Figure 21.4.5-2B: Alternatives to Distribute Overturning Loads 
in Concrete Overlay 

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Techniques for the Seismic Rehabilitation of 
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Figure 21.4.5-2C: Alternatives to Distribute Overturning Loads 
in Concrete Overlay 

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When the joists are perpendicular to the wall, they can be cast into the wall, but when this is 
done, the joists should be treated with preservative, an air gap provided on the top and sides of 
the joists and building paper placed on the bottom to minimize moisture entering the lumber. 
Special rebar detailing will be needed to transfer shear from above the floor to below the floor 
through the weakened area of the joists. Given these issues, the perpendicular joists are often 
headed off and supported off a ledger on the face of the new wall. This requires shoring, but 
simplifies the remaining construction and provides for a better wall. This approach is shown in 
Figure 21.4.5-1. 

Pre-wetting the masonry wall: Cast-in-place walls and, to a lesser extent, shotcrete walls have 
moisture in them during placement that will be absorbed by the masonry. Some engineers 
require wetting the masonry wall, just prior to placement. 

Curing considerations in existing building: Curing concrete emits moisture. If the building has 
finishes that are sensitive to moisture emission, precautions will need to be taken to protect the 
finishes. This can be particularly critical if shotcrete is used. Curing of the face of the concrete 
is either done with curing compound, continuous spraying or a moisture-retaining cover. 
Finding a curing compound that will later be acceptable for certain adhered finishes can be 
difficult. Continuous spraying, however, adds substantial moisture to the interior space. 

Efflorescence concerns from additives and moisture: Like all concrete, overlays can be 
susceptible to alkali salts leaching to the surface, usually leading to white streaks or spots. These 
stains can come from additives within the concrete or from salts within the masonry wall. Use of 
low-alkali concrete is recommended, and additives should be limited to those known not to lead 
to efflorescence. 

Protection of existing masonry substrate: The main advantage of shotcrete is that the existing 
masonry serves as the backside form and forming the front is unnecessary. In typical stone 
masonry and brick masonry situations, the wall will be adequate to serve as the backside form. 
When the wall is particularly thin or in poor condition, the contractor will have to take care to 
brace the masonry to resist the force of the applied shotcrete. 

Cost/Disruption 

Adding new concrete, particularly with shotcrete, can be quite disruptive. Where access is 
sufficient, shotcrete is typically chosen as it is less expensive than cast-in-place work which 
requires front-side formwork. 

Shotcrete: Placing shotcrete requires access for the hose and concrete truck and sufficient room 
(several feet) to spray the concrete. It is desirable to shoot downward, so scaffolding is needed at 
upper portions of walls to achieve the necessary angle. Spraying is noisy and very dusty. If an 
indoor wall is being shot, the room will usually be sealed off with plastic sheeting to control 
dust. During shooting, residueknown as reboundforms at the base of the shoot and must be 
cleaned away so that it does not become part of the overlay. Protection against rebound on 
existing floor and wall surfaces is needed. 

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Cast-in-place concrete: Placing cast-in-place concrete also requires access for the hose and 
concrete truck. Less front-side access is needed than shotcrete, but a front-side form is required 
with the associated sawing and hammering noise of construction. Concrete placement is noisy, 
and in addition to workmen and concrete truck noise, there is the vibrator used to consolidate the 
concrete. 

Proprietary Issues 

There are no known proprietary concerns with shotcrete or cast-in-place overlays on existing 
masonry walls. 

21.4.6 Add Fiber-Reinforced Polymer Overlay to Masonry Wall 
Deficiency Addressed by Rehabilitation Technique 

Improving inadequate in-plane wall strength is the primary purpose of a new fiber-reinforced 
polymer (FRP) overlay, but the overlay can also improve out-of-plane bending capacity. 

Description of the Rehabilitation Technique 

An FRP overlay, typically made of glass or carbon fibers in an adhesive matrix, is applied 
against an existing unreinforced masonry wall to increase the shear strength of the wall. The 
existing wall surface must be prepared to receive the new material, and after application the fiber 
composite must be protected against ultraviolet rays. 

Design Considerations 

Research basis: Research in fiber composites is extensive, but has primarily been focused on 
enhancement of concrete elements or reinforced concrete masonry. There is, nonetheless, a 
growing body of research for unreinforced masonry strengthening. A partial listing of some 
papers is given here. 

For out-of-plane strengthening of unreinforced masonry, tests include Reinhorn and Madan 
(1995a) on clay brick; Portland State University (1998) on hollow clay tile; Vandergrift, 
Gergely, and Young (2002) on hollow concrete masonry; Tumialan, Galati, Namboorimadathil, 
and Nanni (2002) on surface applied fiber reinforced bars to hollow concrete masonry; 
Tumialan, et al. (2002) on glass and aramid fiber reinforced polymer composites on both 
concrete and clay brick; Tumialan, Galati, and Nanni (2002) on situ field tests in an infill frame 
building being demolished of brick and clay tile walls strengthened using glass fiber strips; 
Ehsani, Saadatmanesh, and Velazquez-Dimas (1999); and Velazquez-Dimas, Ehsani, and 
Saadtmanesh (2000) on half-scale clay masonry. 

In-plane testing for clay brick masonry includes Reinhorn and Madan (1995b); Ehsani and 
Saadatmanesh (1996); Ehsani, Saadatmanesh, and Al-Saidy (1997); Haroun and Mosallam 
(2002); Senescu and Mosalam (2004). Elgwady, Lestuzzi and Badoux (2003) performed 
dynamic tests on slender and squat hollow clay masonry piers with and without aramid, glass, 
and carbon fiber composite overlays. Vandergrift, Gergely, and Young (2002) performed tests on 
half-scale hollow concrete masonry. Schwegler and Ketterborn (1996) discusses in-plane 
masonry strengthening with carbon fiber, including straps at various angles of orientation. 
Reinhorn and Madan (1995b) found about a 120% increase in strength from the composite and 

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some additional displacement capacity in a one-cycle reversed cyclic test. After the fiber 
ruptured, however, masonry cracks immediately widened in a brittle manner. Haroun and 
Mosallam (2002) found a minor increase (about 20%) in strength capacity and about a 30% 
increase in displacement capacity. Cracking in the masonry was the ultimate limit which 
occurred after the composite debonded. Senescu and Mosalam (2004) used monotonic diagonal 
tension tests with mixed results, some showing improvements in strength and displacement 
capacity, others actually showing reductions in strength and displacement capacity. Paquette, 
Bruneau, and Brzev (2004) and Paquette and Bruneau (2004) investigated strengthening or 
repairing a one-story building with rocking-critical piers using fiber composite chord strips at 
ends of piers. 

Design basis: There are no code guidelines or FEMA 356 provisions explicitly addressing FRP 
overlays on unreinforced masonry. Information can be obtained from manufacturer literature 
and can be used in conjunction with criteria in the ICC-ES interim standard (ICC-ES, 2003). 
They focus only on the design of the fiber itself, not the fiber and masonry combined 
performance. Velazquez-Dimas and Ehsani (2000) provide modeling and design 
recommendations for out-of-plane strengthening. 

Behavioral mode: It is important to understand the underlying governing behavioral mode of 
both the unstrengthened unreinforced masonry wall and the strengthened wall. Fiber composites 
have little ductility. Adding an FRP overlay to a rocking critical wall pier may be able to reduce 
cyclic degradation of the pier, but will not change the strength or behavior mode if the fiber is 
only applied to the pier. If the fiber crosses the top and bottom of the pier into the spandrel, in 
can inhibit or prevent formation of a rocking mode. Strength will be increased, but ductility will 
be reduced from that of a rocking-critical mode to that of a shear critical mode. 

Detailing and Construction Considerations 

Detailing and construction considerations for FRP overlays include the following. 

Surface preparation: The surface of the masonry needs to be cleaned of loose material and 
finishes that prevent proper adhesion. Sandblasting of the masonry is not usually necessary; a 
wire brush is used instead. See Figure 24.4.6-1. 

Complete overlay vs. strips: Both in research and in practice, both complete overlays over the 
full surface of the wall and use of strips are found. Vertical strips are used when only 
improvement to out-of-plane resistance is needed. Diagonal strips have been used to resist 
diagonal tension stresses from in-plane shear. 

One side or both sides of wall: Applying fiber to both sides of the wall improves performance, 
particularly for out-of-plane resistance, but testing has been performed with fiber on only one 
side. 

Continuity of fiber at top and bottom: Providing load transfer with the fiber can challenging, 
particularly at floor-to-wall interfaces. If the fiber is being used to resist out-of-plane loads and 
is transferring these loads back into the floor diaphragm, special details may be needed to turn 
the vertical fiber overlay into the horizontal diaphragm. Fiber cannot be bent at 90 degrees; 

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Techniques for the Seismic Rehabilitation of 
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Figure 21.4.6-1: Fiber Composite Wall Overlay on URM Wall 

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rather a radius is needed. Sometimes, steel reinforcing plates are used to stiffen the turns. If the 
fiber is being used to transfer in-plane loads from one story to the next, continuity past the floor 
is needed and will require special details as does shear transfer out of the diaphragm into the wall 

Moisture barrier: Fiber composites are impermeable. If continuous overlays are used, moisture 
transmission through the masonry wall will be stopped at the fiber. Eventually, the concern 
would be the moisture would build up and begin to delaminate the fiber bond and lead to general 
building concerns with excessive moisture. 

Additional information: See Section 13.4.1 for more detailed discussion on FRP issues including: 

Composite makeup and application 

Mechanical properties 

Fiber and mechanical anchors 

Durability 

Constructability 

Cost/Disruption 

Fiber composites are relatively expensive as an application for masonry wall strengthening and 
have not seen significant use. Disruption comes from the sandblasting and surface preparation of 
the masonry wall, the fumes from the adhesives used in application of the fiber composite and 
removal and access requirements where the fiber transitions from story to story at the floor levels. 

Proprietary Issues 

Fiber composite materials are supplied by vendors. Capacities and design methods vary 
depending on the vendors. 

21.4.7 Infill Opening in a URM Wall 
Deficiency Addressed by Rehabilitation Technique 

Inadequate unreinforced masonry in-plane wall strength. 

Description of the Rehabilitation Technique 

Window and door openings are filled to increase the shear capacity and reduce the shear stresses 
on the unreinforced masonry wall. The opening is typically filled with concrete, reinforced 
concrete masonry units, or reinforced clay brick, rather than with unreinforced masonry due to 
code concerns with adding unreinforced masonry. To provide adequate shear transfer between 
the existing wall and the new infill, the interface can be toothed, but more typically, drilled 
dowels are used. See Figure 21.4.7-1. 

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Figure 21.4.7-1: Infilling an Opening in a URM Wall 

Design Considerations 

Research basis: Research that directly addresses testing of infilled openings has not been 
identified. 

Capacity of infill: The typical infill materials are stronger and potentially stiffer than the 
surrounding masonry. It is typical, however, to consider the infilled wall as solid unreinforced 
masonry in determining shear capacity of the composite wall. Since the increase in shear 
capacity from the infill is often not substantial, it is done when only moderate increases in 
capacity are needed. 

Behavioral mode: It is important to understand the underlying governing behavioral mode of 
both the unstrengthened unreinforced masonry wall and the infilled wall. Infilling openings 
could change a rocking-critical wall line to a shear-critical wall line, which may be less 
desirable. 

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Detailing and Construction Considerations 

Aethestics: Infilling openings can obviously have a significant visual impact. Sometimes new 
concrete is used and is set back from the exterior face so that a window can be placed. Lighting 
can be added between the glazing and infill to mitigate the opacity of the infill. 

Cost/Disruption 

Infilling an opening is relatively inexpensive if no architectural treatment is done to the face. 
Disruption is also more localized compared to other in-plane wall strengthening methods like 
concrete and fiber reinforced polymer overlays. 

Noise will occur during drilling holes for drilled dowels and placing the infill. 

Proprietary Issues 

There are no proprietary concerns with infilling masonry wall openings. 

21.4.8 Add Concrete or Masonry Shear Wall (Connected to a Wood Diaphragm) 
Deficiency Addressed by Rehabilitation Technique 

A new concrete or masonry wall provides additional global strength and stiffness, reduces 
demands on existing masonry walls and can reduce demands on diaphragms by cutting tributary 
spans. 

Description of the Rehabilitation Technique 

The new wall should be properly designed to meet current code detailing provisions. This 
section focuses on detailing at the interface between the new wall and the existing wood 
diaphragm. Figure 21.4.8-1 shows sample concepts both for joists parallel to the wall and joists 
perpendicular to the wall at an interior floor location; Figure 21.4.8-2 shows similar concepts at a 
roof. 

Design Considerations 

Research basis: Research specific to ledger connections between wood diaphragms and concrete 
or masonry walls has not been identified. 

Capacity: The connection can be designed either for the code level demands or to develop the 
diaphragm, depending on where inelastic action is intended to occur. 

Detailing and Construction Considerations 

Detailing and construction considerations for connecting a new wall to an existing wood 
diaphragm include the following. 

Masonry vs. concrete: Masonry is usually considered quicker to install and less expensive; 
concrete (or shotcrete) is stronger and stiffer and usually considered to have better earthquake 
performance. Making connections from concrete to wood diaphragms can be easier than with 
masonry. 

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Figure 21.4.8-1: Connecting a New Concrete Wall to an Existing Wood Floor 

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Figure 21.4.8-2: Connecting a Concrete Wall to an Existing Wood Roof at Interior 

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In-plane shear transfer: In Figure 21.4.8-1, shear transfer from the diaphragm to the wall goes 
from the diaphragm boundary nailing to the ledger and through the threaded rod into the wall. A 
tight fit on the rod and ledger is needed. The ledger should be dry dimensional lumber or glulam 
material to minimize vertical shrinkage of the ledger. When the wall is not as long as the 
diaphragm (a very common occurrence), a collector attachment into the wall will be needed. 
Figure 21.4.8-1A shows a steel angle with headed studs cast into the wall and diaphragm-tocollector 
connections using lag screws. The steel could go above or below the floor. When 
loads are relatively low, wood members such as the ledger can be used as the collector. 

Out-of-plane tension transfer: In Figure 21.4.8-1, tension transfer of wall loads goes into the tie-
down anchor, into the blocking, through straps in the blocking to additional blocks as required 
and eventually back into the diaphragm. Alternatively, blocking for a bay or two can be placed 
on both sides and out-of-plane resistance accomplished by compression bearing on the 
diaphragm joists. 

Joist direction: When the wall can be fit in between existing joists, the amount of labor is 
reduced. When joists are perpendicular to the wall, the joists are typically headed off on each 
side of the wall to allow the wall to pass through. This requires temporary shoring of the floor 
around the wall. At the top of the wall, the wall can stop just under the joists and be blocked up 
to the diaphragm for shear transfer. 

Shotcrete vs. cast-in-place concrete: See Section 21.4.5 for discussion of shotcrete vs. cast-inplace 
concrete issues. 

Cost/Disruption 

See Section 21.4.5 for discussion of cost and disruption issues. 

Proprietary Issues 

There are no proprietary concerns with connecting a concrete or masonry wall to a wood 
diaphragm. 

21.4.9 Add Steel Moment Frame (Connected to a Wood Diaphragm) 
Deficiency Addressed by Rehabilitation Technique 

A new moment frame provides additional global strength, reduces demands on existing masonry 
walls and can reduce demands on diaphragms by cutting tributary spans. 

Description of the Rehabilitation Technique 

When a moment frame is added into a URM building, it typically goes either just behind a 
highly punctured street front faade or at an interior location within the diaphragm. Figure 
21.4.9-1 shows the perimeter condition; Figure 21.4.9-2 shows interior conditions. A moment 
frame retrofit at a W1A building with a soft story is discussed in Chapter 6. 

Design Considerations 

Research basis: New steel moment frame issues are covered by FEMA 350 (FEMA, 2000). The 
CUREE woodframe project report on tuckunder building testing (Mosalam, et al., 2002) 

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documents quasistatic component testing of moment frame to wood diaphragm connections and 
full-scale testing of a three-story tuckunder apartment building rehabilitated with a ground story 
moment frame on the open front side. 

Figure 21.4.9-1: New Perimeter Steel Moment Frame to an Existing Wood Floor 

Stiffness considerations: At either the perimeter or interior condition, reasonable stiffness of the 
frame is desirable. At the perimeter, minimizing the amount of drift and resulting masonry 
faade cracking is desirable. At the interior, if the moment frame does not have sufficient 
stiffness, the diaphragm will span between the end walls with the moment frame taking out 
relatively small loads due to its flexibility. 

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Figure 21.4.9-2: New Interior Steel Moment Frame to an Existing Wood Floor 

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Design forces: The new moment frame design can be governed by either stiffness or strength. 
Strength demands can either be minimum design loads or in some cases the moment frame can 
be designed to be stronger than the diaphragm so inelastic action happens in the diaphragm. For 
connection design of the frame to the diaphragm, it is particularly desirable to make sure the 
connections are stronger than the weaker of the diaphragm or the moment frame. 

Pinned base: To minimize foundation demand requirements, new moment frames in retrofits are 
often designed with pinned bases. 

Detailing and Construction Considerations 

Detailing and construction considerations for connecting a new moment frame to an existing 
wood diaphragm include the following. 

Welding vs. bolting: Welding adjacent to wood framing poses a very real fire hazard. 
Specifications and common sense usually dictate various fire watch provisions in these 
situations. Cases of hot welding slag lost from view and later reigniting wood material after the 
welding for the day was finished have been observed and are particularly troublesome. Where 
possible, detailing with shop welded connections, and then field bolting, is desirable. See 
Chapter 8 for additional comments on welding. 

Connecting directly to the masonry: In Figure 21.4.9-1, the moment frame is connected to both 
the masonry faade and the diaphragm to take out load from the punctured wall into the frame 
and from the diaphragm into the frame. In alternative details, the load can be taken from the wall 
into the diaphragm and then through the diaphragm to the frame. 

Cost/Disruption 

Installation of a new moment frame can be fairly disruptive, though it is usually less disruptive 
than a new wall. The frame is chosen when existing window or door openings need to be 
preserved, but head height and visual issues must be considered. Adding new structural steel 
members can be comparatively expensive, but if the choice is to provide a wood structural panel 
overlay on a floor or add a new moment frame, the new moment frame can often be less 
expensive. 

Proprietary Issues 

There are no proprietary concerns with connecting a steel moment frame to a wood diaphragm. 
Certain moment frame beam-to-column connections may have proprietary considerations. See 
Chapter 8. 

21.4.10 Add or Enhance Crosswalls 
Deficiency Addressed by Rehabilitation Technique 

Inadequate diaphragm strength and/or excessive diaphragm displacement. 

Description of the Rehabilitation Technique 

The ABK research program (ABK, 1984) showed that partition walls, called crosswalls, serve as 
energy-absorbing, displacement-limiting damping elements during seismic loading. The 2003 

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IEBC and 1997 UCBC permit certain qualifying buildings to use the Special Procedure with 
crosswalls as an integral part of the procedure. There are three basic types of crosswalls: 
existing partitions with various sheathing materials, new partitions, and new steel moment 
frames. Existing partitions may be adequate without any rehabilitation or they may need 
strengthening if they are not connected to the diaphragms or have insufficient capacity. New 
wood structural panel partitions must also be connected to the diaphragm and meet certain 
minimum capacities. New moment frames must meet minimum strength and stiffness criteria. 
See Figures 21.4.10-1 to 21.4.10-3. 

Figure 21.4.10-1: Strengthen Existing Crosswalls 

Design Considerations 

Research basis: ABK (1984) provides background for the basis of the crosswall concept. 

Qualifying buildings: In the 2003 ICBC, the Special Procedure can be used only with buildings 
have flexible diaphragms at all levels and meet certain requirements regarding open fronts and 
number of wall lines in each direction. For the 1997 UCBC, these requirements apply, and 

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buildings must be a maximum of six stories, and they cannot be essential or hazardous facilities. 
The crosswall concept and the Special Procedure were not adopted in FEMA 356. 

Figure 21.4.10-2: Add New Crosswalls 

Crosswall requirements: The 1997 UCBC and the 2003 IEBC have a number of requirements 
on crosswall locations, aspect ratios, connection strength, spacing limits that need to be satisfied. 

New crosswalls: New crosswalls will typically be done with structural wood panels and are 
similar to adding new wood structural panel shear walls. See Chapters 5 and 6 for additional 
information. 

Steel moment frames: Adding moment frames for use as crosswalls is very similar to adding 
moment frames for use as new lateral force-resisting elements. See Section 21.4.9 for additional 
information. 

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Figure 21.4.10-3: Add New Moment Frame as Crosswall 

Detailing and Construction Considerations, and Cost/Disruption 

Adding or enhancing woodframe crosswalls is similar to adding or enhancing woodframe shear 
walls; see Chapters 5 and 6. Adding a moment frame as a crosswall is similar to adding a 
moment frame as a lateral force-resisting element; see Section 21.4.9. 

Proprietary Issues 

There are no proprietary concerns with adding crosswalls. Certain steel moment frame beam-tocolumn 
connections are proprietary. 

21.4.11 Add Supplemental Vertical Support for Truss or Girder 
Deficiency Addressed by Rehabilitation Technique 

Supplemental vertical supports provide a secondary load path for concentrated gravity loads on 
unreinforced masonry walls. 

Description of the Rehabilitation Technique 

A steel or wood post is added under existing trusses and girders. 

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Design Considerations 

Research basis: There are no known tests of supplemental supports. 

Purpose of the support: To some engineers, the goal of adding a supplemental support is to 
provide a back-up gravity load path if there is local deterioration of the masonry underneath a 
concentrated load like a truss bearing point. To others, it provides support if more wholesale 
failure of the wall occurs. 

Independence of the support: The 1997 UCBC and 2003 IEBC both use the term independent 
secondary columns when referring to supplemental vertical supports. To some engineers, 
independent means separated from the wall. A post an inch or two away from the wall 
satisfies this requirement. To other engineers, independent simply means an alternative 
support, so that a ledger or pilaster on the wall is sufficient. In this scenario, the wall just 
beyond the damaged area is assumed to remain intact enough that gravity load resistance is not 
compromised. In Figure 21.4.11-1, a gap is shown. 

Use as out-of-plane brace also: Some engineers like to take advantage of supplemental vertical 
support posts to serve as out-of-plane braces for the walls as well. Combined bending and axial 
demands must be considered. 

Continuity of support: The 1997 UCBC and 2003 IEBC both do not specify whether the posts 
need to continue down to the next story. It is common to transfer loads at the base of upper 
story supplemental supports back to existing framing. This framing must be adequate to take the 
loads that would occur if they supplemental support began to take load. 

Foundation support for posts: The 1997 UCBC and 2003 IEBC both do not explicitly specify 
whether the posts need a compliant new foundation or whether the posts can simply bear on an 
existing slab-on-grade. SEAOC (1992), however, states that a foundation is not required for the 
posts. Nonetheless, some engineers believe it is prudent to check the slab for bearing support 
and provide additional support if needed. 

Triggering elements: The 1997 UCBC and 2003 IEBC only indicate that triggering elements are 
trusses and beams, others than rafters and joists. The implication is that elements that support 
other structural members are the primary focus. 

Detailing and Construction Considerations 

Detailing and construction considerations for supplemental vertical supports include the 
following. 

Steel vs wood: Supplemental support posts can either be of steel or wood. Steel members are 
smaller; wood members are less expensive. 

Finish the new elements are leave bare: For certain architectural approaches, leaving the 
supplemental posts bare is compatible with the existing aesthetic. If it is not, the posts can be 
furred at added cost. 

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Figure 21.4.11-1: Supplemental Vertical Support 

Cost/Disruption 

The relative cost of adding supplemental supports depends on the number used, whether they 
continue down to the ground and whether a new foundation is installed. Interior occupants will 
be disrupted locally as the posts are installed, and the usable space in the vicinity of the posts 
will be reduced. 

Proprietary Considerations 

There are no known proprietary concerns with employing supplemental vertical supports. 

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21.4.12 Add Veneer Ties in a URM Wall 
Deficiency Addressed by Rehabilitation Technique 

Missing or inadequate ties between a masonry veneer wythe and the backing wythes can lead to 
delamination of the veneer and a falling hazard. 

Description of the Rehabilitation Technique 

If the front or facing wythe of brick is not integrally tied into the interior wythes with header 
courses, it is considered a veneer. If sufficient metal veneer ties are not present to anchor the 
veneer to the backing wythe, new ties can be provided. Figure 21.4.12-1 shows anchorage using 
drilled dowels to connect the wythes, either from the exterior or interior, in a typical brick wall. 
Figure 21.4.12-2 shows anchorage between stone facing and brick backing. 

Figure 21.4.12-1: Veneer Ties in Brick Masonry 

Design Considerations 

Research basis: No academic research on veneer ties has been identified. Individual vendors 
have performed internal testing of their own products. 

Veneer definition: Codes such as the 2003 ICBC and 1997 UCBC give minimum lay-up 
requirements for multi-wythe solid brick. The facing and backing wythes are to be bonded so 
than not less than 10 percent of the exposed face area is composed of solid headers extending 
less than 4 inches into the backing. The clear distance between adjacent full-length headers shall 
not exceed 24 inches vertically or horizontally. Facing wythes that do not meet these lay-up 

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requirements are to be considered as veneer. Veneer wythes are not used in resisting shear 
forces in the wall and do not count in the thickness used for determining out-of-plane bending 
resistance. 

Figure 21.4.12-2: Veneer Ties for Stone Masonry Facing 

Veneer tie requirements: Codes such as the 2003 ICBC and 1997 UCBC provide acceptance 
criteria for existing ties. They are to be corrugated galvanized iron strips shown to be in good 
conditions, with dimensions no less than 1 wide, 8 long and 1/16 thick and with a maximum 
spacing of 24 on center and a maximum supported area of four square feet. Veneer ties not 
meeting these requirements are to be strengthened. 

Detailing and Construction Considerations 

Detailing and construction considerations for veneer ties include the following. 

Types of ties: The ties used in new construction are typically inappropriate for rehabilitation 
since they are installed as the wall is built up. There are, however, a fair number of proprietary 
products made by masonry accessory manufacturers that can be used in retrofit applications. 
Some involve expansion anchors in the backing wythes. Others involve helical anchors or spiral 
ties that screw into the backing wythes. Figures 21.4.12-1 and 21.4.12-2 show traditional 
drilled dowels. When drilled dowels are used, the tie diameter does not need to be large since 
loads are low and minimizing the size of the hole is important. 

Drilled dowels installation face: Drilled dowels can be installed from the interior or exterior 
depending on which face is more sensitive. 

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Location of drilled dowel: The most reliable location of the drilled dowel is in the center of the 
brick, but this is will cause the largest aesthetic impact. The dowel can be placed in the bed joist 
or bed and head joint intersection to minimize the impact. Recessing the tip of the dowel and 
covering the end with repointing mortar is recommended. 

Brick veneer vs. stone veneer: Anchoring thick stone is usually easier to do from the interior 
because the thickness of the stone permits greater cover on the face, reducing the likelihood of 
spalling. With random ashlar layout and interior installation, however, trying to locate the dowel 
away from the edges of the stone is not practical. 

Corrosion considerations: Any dowel installed from the exterior should be done in stainless 
steel to minimize corrosion. 

Cost and Disruption Considerations 

Veneer anchorage can be relatively expensive depending on the number of new ties added. Ties 
installed from the inside are much more disruptive. See Section 21.4.2 for additional 
information on drilled dowels. 

Proprietary Considerations 

Many of the veneer tie anchorage systems are proprietary. 

21.5 References 
ABK, 1981a, Methodology for Mitigation of Seismic Hazards in Existing Unreinforced Masonry 
Buildings: Categorization of Buildings, a joint venture of Agbabian Associates, S.B. Barnes and 
Associates, and Kariotis and Associates (ABK), Topical Report 01, c/o Agbabian Associates, El 
Segundo, CA. 

ABK, 1981b, Methodology for Mitigation of Seismic Hazards in Existing Unreinforced Masonry 
Buildings: Diaphragm Testing, a joint venture of Agbabian Associates, S.B. Barnes and 
Associates, and Kariotis and Associates (ABK), Topical Report 03, c/o Agbabian Associates, El 
Segundo, CA. 

ABK, 1981c, Methodology for Mitigation of Seismic Hazards in Existing Unreinforced Masonry 
Buildings: Wall Testing (Out-of-Plane), a joint venture of Agbabian Associates, S.B. Barnes and 
Associates, and Kariotis and Associates (ABK), Topical Report 04, c/o Agbabian Associates, El 
Segundo, CA. 

ABK, 1984, Methodology for Mitigation of Seismic Hazards in Existing Unreinforced Masonry 
Buildings: The Methodology, a joint venture of Agbabian Associates, S.B. Barnes and 
Associates, and Kariotis and Associates (ABK), Topical Report 08, c/o Agbabian Associates, El 
Segundo, CA. 

Abrams, D.P. and J.M Lynch, 2001, Flexural Behavior of Retrofitted Masonry Piers, EERCMAE 
Joint Seminar on Risk Mitigation for Regions of Moderate Seismicity, IL. 

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Breiholtz, D., 1987, Centercore Strengthening System for Seismic Hazard Reduction of 
Unreinforced Masonry Bearing Wall Buildings, Proceedings of the 56th Annual Convention of 
the Structural Engineers Association of California, San Diego, CA. 

Breiholz, D.C., 1992, Center Core Seismic Hazard Reduction System for URM Buildings, 
Proceedings of the Tenth World Conference on Earthquake Engineering, Vol. 9, pp. 5395-5399. 

Ehsani, M.R. and H. Saadatmanesh, 1996, Seismic Retrofit of URM Walls with Fiber 
Composites, The Masonry Society Journal, Spring. 

Ehsani, M.R. H. Saadatmanesh, and A. Al-Saidy, 1997, Shear Behavior of URM Retrofitted 
with FRP Overlays, Journal of Composites for Construction, ASCE 1(1), 17-25. 

Ehsani, M.R. H. Saadatmanesh, and J.I. Velazquez-Dimas, 1999, Behavior of Retrofitted URM 
Walls Under Simulated Earthquake Loading, Journal of Composites for Construction, ASCE 
4(3), 134-142. 

ElGawady, M., P. Lestuzzi, and M. Badoux, 2003, Rehabilitation of Unreinforced Brick 
Masonry Walls Using Composites, Architectural and Structural Design of Masonry With Focus 
on Retrofitting of Masonry Structures and Earthquake Resistant Design, International Short 
Course, Dresden University of Technology, December 7-18, Dresden, Germany. 

ElGawady, M, P. Lestuzzi, and M. Badoux, 2004, A Review of Conventional Seismic 
Retrofitting Techniques for URM, 13th International Brick and Block Masonry Conference, 
Amsterdam, July 4-7. 

FEMA, 1997a, NEHRP Guidelines of the Seismic Rehabilitation of Buildings, FEMA 273, 
Federal Emergency Management Agency, Washington, D.C., October. 

FEMA, 1997b, NEHRP Commentary on the Guidelines of the Seismic Rehabilitation of 
Buildings, FEMA 274, Federal Emergency Management Agency, Washington, D.C., October. 

FEMA, 1999a, Evaluation of Earthquake Damaged Concrete and Masonry Wall Buildings: 
Basic Procedures Manual, FEMA 306, May. 

FEMA, 1999b, Evaluation of Earthquake Damaged Concrete and Masonry Wall Buildings: 
Technical Resources, FEMA 307, May. 

FEMA, 2000, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 
356, Federal Emergency Management Agency, Washington, D.C., November. 

Ganz, H.R., 1993, Strengthening of Masonry Structures with Post-Tensioning, Proceedings of 
the Sixth North American Masonry Conference, Philadelphia, Pennsylvania, June 6-9, pp. 
645-655. 

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Ghanem, G., M.A. Zied and A.E. Salama, 1994, Retrofit and Strengthening of Masonry 
Assemblages Using Fiberglass, Proceedings of the Tenth International Brick and Block 
Masonry Conference, Vol. 2, pp. 499-508. 

Haroun, M.A and A.S Mosallam, 2002, Cyclic Shear Test of Multi-Wythe Existing Brick Wall: 
Retrofitted by a Single Layer of TYFO SHE-51A Applied on One Side, University of California, 
Irvine and California State University Fresno, February. 

Hutchison, D.L., P.M.F. Yong and G.H.F. McKenzie, 1984, Laboratory Testing of a Variety of 
Strengthening Solutions for Brick Masonry Wall Panels, Proceedings of the Eighth World 
Conference on Earthquake Engineering, Vol. 1, pp. 575-582. 

ICBO, 1997, Uniform Code for Building Conservation, 1997 Edition, International Conference 
of Building Officials, Whittier, CA. 

ICC, 2003, International Existing Building Code, 2003 Edition, International Code Council, 
Country Club Hills, IL. 

ICC-ES, 2003, Interim Criteria for Concrete and Reinforced and Unreinforced Masonry 
Strengthening Using Fiber-Reinforced Polymer (FRP) Composite Systems, Acceptance Criteria 
AC-125, June, International Code Council Evaluation Service, Whittier, CA. 

ICC-ES, 2005, Acceptance Criteria for Anchors in Unreinforced Masonry Elements (AC60), 
Approved April 2005, Whittier, CA. 

Jurukovski, D., L. Krstevska, R. Alessi, P.P. Diotallevi, M. Merli and F. Zarri, 1992, Shaking 
Table Tests of Three Four-Storey Brick Masonry Models: Original and Strengthened by RC 
Core and RC Jackets, Proceedings of the Tenth World Conference on Earthquake Engineering, 
July, Madrid, A A Balkema, Rotterdam, Vol. 5, pp. 2795-2800. 

Kahn, L.F., 1984, Shotcrete Retrofit of Unreinforced Brick Masonry, Proceedings of the 
Eighth World Conference on Earthquake Engineering, San Francisco, CA, July, pp. 583-590. 

Kariotis, J., 1982, Bracing of Unreinforced Masonry Walls, Out-of-Plane, Earthquake Hazard 
Mitigation of Unreinforced Masonry Buildings Built Prior to 1934, Seminar Proceedings, 
Structural Engineers Association of Southern California, April, Whittier, CA. 

Mosalam, K., et al., 2002, Seismic Evaluation of an Asymmetric Three-Story Woodframe 
Building, CUREE Publication No. W-19, Consortium of Universities for Research in Earthquake 
Engineering, Richmond, CA. 

Paquette, J., M. Bruneau, and S. Brzev, 2003, Pseudo-Dynamic Testing of Unreinforced 
Masonry Building with Flexible Diaphragm, Journal of Structural Engineering, Vol. 129, No. 
6, June, pp. 708-716. 

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Paquette, J., M. Bruneau, and S. Brzev, 2004, Seismic Testing of Repaired Unreinforced 
Masonry Building Having Flexible Diaphragm, Journal of Structural Engineering, Vol. 130, 
No. 10, October, pp. 1487-1496. 

Paquette, J. and M. Bruneau, 2004, Pseudo-Dynamic Testing of Unreinforced Masonry 
Building with Flexible Diaphragm, Proceedings of the 13th World Conference on Earthquake 
Engineering, Vancouver, B.C., Paper 2609. 

Plecnik, J., T. Cousins and E. OConner, 1986, Strengthening of Unreinforced Masonry 
Buildings, Journal of Structural Engineering, American Society of Civil Engineers, Vol. 112, 
No. 5, May, pp. 1070-1087. 

Plecnik, J.M., 1988, Preliminary Report on Summary of the Second Meeting of the Advisory 
Committee for NSF Research Project on the Centercore Rehabilitation Technique, Structures 
Laboratory Report 88-8-18, California State University, Long Beach, August 17. 

Portland State University, 1998, Full Scale Tests of Retrofitted Hollow Clay Tile Walls, Final 
Report Prepared for Contech, Department of Civil Engineering, June. 

Reinhorn, A.M. and A. Madan, 1995a, Evaluation of Tyfo-W Fiber Wrap System for Out of 
Plane Strengthening of Masonry Walls, Report No. AMR 95-2001, Department of Civil 
Engineering, State University of New York at Buffalo, Buffalo, NY, March. 

Reinhorn, A.M. and A. Madan, 1995b, Evaluation of Tyfo-W Fiber Wrap System for In Plane 
Strengthening of Masonry Walls, Report No. AMR 95-2002, Department of Civil Engineering, 
State University of New York at Buffalo, Buffalo, NY, August. 

Rosenboom, O.A. and M.J. Kowalsky, 2004, Reversed In-Plane Cyclic Behavior of 
Posttensioned Clay Brick Masonry Walls, Journal of Structural Engineering, Vol. 130, No. 5, 
May, pp. 787-798. 

Rutherford & Chekene, 1990, Seismic Retrofitting Alternatives for San Franciscos Unreinforced 
Masonry Buildings: Estimates of Construction Cost & Seismic Damage, for the Department of 
City Planning of the City and County of San Francisco, Oakland, CA, May. 

Rutherford & Chekene, 1997, Development of Procedures to Enhance the Performance of 
Rehabilitated Buildings, prepared by Rutherford & Chekene Consulting Engineers, published by 
the National Institute of Standards and Technology as Reports NIST GCR 97-724-1 and NIST 
97-724-2. 

Schwegler, G. and P. Kelterborn, 1996, Earthquake Resistance of Masonry Structures 
Strengthened with Fiber Composites, Proceedings of Eleventh World Conference on 
Earthquake Engineering, Acapulco, June, Elsevier Science Ltd., Paper No. 1460. 

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Senescu, R. and K.M. Mosalam, 2004, Retrofitting of Unreinforced Masonry Walls Using Fiber 
Reinforced Polymer Laminates, Report UCB/SEMM 2004/3, University of California, Berkeley, 
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SEAOC (Structural Engineers Association of California), 1992, Commentary on Appendix 
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