Building Seismic Safety Council
of the National Institute of Building Sciences 

Program on 
Improved Seismic Safety 
Provisions 

2003 Edition 

RECOMMENDED PROVISIONS 
FOR SEISMIC REGULATIONS 
FOR NEW BUILDINGS 
AND OTHER STRUCTURES (FEMA 450) 

Part 2: Commentary 


 The Building Seismic Safety Council (BSSC) was established in 1979 under the auspices of the 
National Institute of Building Sciences as an entirely new type of instrument for dealing with the 
complex regulatory, technical, social, and economic issues involved in developing and promulgating 
building earthquake hazard mitigation regulatory provisions that are national in scope. By bringing 
together in the BSSC all of the needed expertise and all relevant public and private interests, it was 
believed that issues related to the seismic safety of the built environment could be resolved and 
jurisdictional problems overcome through authoritative guidance and assistance backed by a broad 
consensus. 

The BSSC is an independent, voluntary membership body representing a wide variety of building 
community interests. Its fundamental purpose is to enhance public safety by providing a national forum 
that fosters improved seismic safety provisions for use by the building community in the planning, 
design, construction, regulation, and utilization of buildings. 
See the back of this Commentary volume for a full description of BSSC activities. 

2003 BSSC BOARD OF DIRECTION 

Chairman 
Charles Thornton, Chairman/Principal, Thornton-Tomasetti Group, Inc., New York, New 
York 

Vice Chairman 
David Bonneville, Degenkolb Engineers, San Francisco, California 

Secretary 
Charles Carter, Chief Structural Engineer, American Institute of Steel Construction, Chicago, 
Illinois 

Ex-Officio 
William W. Stewart, Stewart-Schaberg Architects, Clayton, Missouri (representing the 
American Institute of Architects) 

Members 
J. Gregg Borchelt, Vice President, Brick Industry Association, Reston, Virginia 
Edwin Dean, Nishkian Dean, Portland, Oregon 
Bradford K. Douglas, Director of Engineering, American Forest and Paper Association, 
Washington, D.C. 
Henry Green, Executive Director, Bureau of Construction Codes and Fire Safety, State of 
Michigan, Department of Labor and Economic Growth, Lansing, Michigan (representing the 
National Institute of Building Sciences) 
H.S. Lew, Senior Research Engineer, National Institute of Standards and Technology, 
Gaithersburg, Maryland (representing Interagency Committee on Seismic Safety in 
Construction) 
Joseph Messersmith, Coordinating Manager, Regional Code Services, Portland Cement 
Association, Rockville, Virginia (representing the Portland Cement Association) 
Jim Rinner, Project Manager II, Kitchell CEM, Sacramento, California 
James Rossberg, Manager, Technical Activities for the Structural Engineering Institute, 
American Society of Civil Engineers, Reston Virginia 
Jeffery Sciadone, Associate Director, of Engineering, Institute of Business and Home Safety, 
Tampa, Florida 
W. Lee Shoemaker, Director, Engineering and Research, Metal Building Manufacturers 
Association, Cleveland, Ohio 
Howard Simpson, Simpson Gumpertz and Heger, Arlington, Massachusetts (representing 
National Council of Structural Engineers Associations) 
Charles A. Spitz, Architect/Planner/Code Consultant, Wall New Jersey (representing the 
American Institute of Architects) 

BSSC STAFF 
Claret M. Heider, Vice President for BSSC Programs 
Bernard F. Murphy, Director, Special Projects 
Carita Tanner, Communications/Public Relations Manager 
Patricia Blasi, Administrative Assistant 

 BSSC Program on Improved Seismic Safety Provisions 

NEHRP RECOMMENDED PROVISIONS 
(National Earthquake Hazards Reduction Program) 

FOR SEISMIC REGULATIONS 

FOR NEW BUILDINGS AND 
OTHER STRUCTURES (FEMA 450) 

2003 EDITION 

Part 2: COMMENTARY 

Prepared by the 
Building Seismic Safety Council 
for the 
Federal Emergency Management Agency 

BUILDING SEISMIC SAFETY COUNCIL 
NATIONAL INSTITUTE OF BUILDING SCIENCES 
Washington, D.C. 
2004 
 ii 
NOTICE: Any opinions, findings, conclusions, or recommendations expressed in this publication do 
not necessarily reflect the views of the Federal Emergency Management Agency. Additionally, neither 
FEMA nor any of its employees make any warranty, expressed or implied, nor assume any legal liability 
or responsibility for the accuracy, completeness, or usefulness of any information, product, or process 
included in this publication. 
This report was prepared under Contract EMW-2001-CO-0269 
between the Federal Emergency Management Agency and the 
National Institute of Building Sciences. 
Building Seismic Safety Council activities and products are 
described at the end of this report. For further information, see the 
Council website (www.bssconline.org) or contact the Building 
Seismic Safety Council, 1090 Vermont, Avenue, N.W., Suite 700, 
Washington, D.C. 20005; phone 202-289-7800; fax 202-289-1092; 
e-mail bssc@nibs.org. 
Copies of this report on CD Rom may be obtained from the FEMA 
Publication Distribution Facility at 1-800-480-2520. Limited paper 
copies also will be available. The report can also be downloaded in 
pdf form from the BSSC website at www.bssconline.org. 
The National Institute of Building Sciences and its Building Seismic 
Safety Council caution users of these Provisions documents to be 
alert to patent and copyright concerns especially when applying 
prescriptive requirements. 
 iii 
CONTENTS 
Chapter 1 GENERAL PROVISIONS ....................................................................................1 
1.1 GENERAL ....................................................................................................................1 
1.1.1 Purpose..........................................................................................................1 
1.2 SEISMIC USE GROUPS..............................................................................................4 
1.2.5 Seismic Use group III structure access protection ........................................8 
1.3 OCCUPANCY IMPORTANCE FACTOR...................................................................8 
1.4 SEISMIC DESIGN CATEGORY.................................................................................8 
1.4.2 Site limitation for Seismic Design Categories E and F ...............................10 
1.5 SEISMIC DESIGN CATEGORY A...........................................................................10 
1.5.1 Lateral forces ..............................................................................................10 
1.5.2 Connections.................................................................................................11 
1.5.3 Anchorage of concrete or masonry walls ....................................................11 
Chapter 2 QUALITY ASSURANCE...................................................................................13 
2.1 GENERAL .................................................................................................................13 
2.1.1 Scope ...........................................................................................................13 
2.2 GENERAL REQUIREMENTS ..................................................................................14 
2.3 SPECIAL INSPECTION ............................................................................................15 
2.3.9 Architectural components ...........................................................................15 
2.3.10 Mechanical and electrical components......................................................15 
2.4 TESTING....................................................................................................................16 
2.4.5 Mechanical and electrical equipment ..........................................................16 
2.5 STRUCTURAL OBSERVATIONS ...........................................................................16 
2.6 REPORTING AND COMPLIANCE PROCEDURES...............................................16 
Chapter 3 GROUND MOTION ...........................................................................................17 
3.1 GENERAL ..................................................................................................................17 
3.1.3 Definitions...................................................................................................17 
3.2 GENERAL REQUIREMENTS ..................................................................................18 
3.2.2 Procedure selection .....................................................................................18 
3.3 GENERAL PROCEDURE..........................................................................................18 
3.3.2 Site coefficients and adjusted acceleration parameters ...............................19 
3.3.4 Design response spectrum...........................................................................26 
3.4 SITE SPECIFIC PROCEDURE..................................................................................27 
3.4.2 Deterministic maximum considered earthquake .........................................29 
3.5 SITE CLASSIFICATION FOR SEISMIC DESIGN..................................................29 
3.5.1 Site class definitions....................................................................................29 
3.5.2 Steps for classifying a site...........................................................................30 
Chapter 4 STRUCTURAL DESIGN CRITERIA ................................................................35 
4.1 GENERAL ..................................................................................................................35 
4.1.2 References ..................................................................................................35 
4.2 GENERAL REQUIREMENTS ..................................................................................35 
4.2.1 Design basis ................................................................................................35 
4.2.2 Combination of load effects ........................................................................39 
4.3 SEISMIC-FORCE-RESISTING SYSTEM ................................................................44 
4.3.1 Selection and limitations .............................................................................44 
4.3.2 Configuration ..............................................................................................47 
4.3.3 Redundancy.................................................................................................51 
 iv 
4.4 STRUCTURAL ANALYSIS...................................................................................... 52 
4.4.1 Procedure selection ..................................................................................... 52 
4.4.2 Application of loading ................................................................................ 52 
4.5 DEFORMATION REQUIREMENTS........................................................................ 52 
4.5.1 Deflection and drift limits .......................................................................... 52 
4.5.3 Seismic Design Categories D, E, and F ...................................................... 54 
4.6 DESIGN AND DETAILING REQUIREMENTS ..................................................... 55 
4.6.1 Seismic design Category B ......................................................................... 55 
4.6.2 Seismic design Category C ......................................................................... 58 
4.6.3 Seismic Design Category D, E, and F......................................................... 58 
ALTERNATIVE SIMPLIFIED CHAPTER 4..................................................................... 59 
Chapter 5 STRUCTURAL ANALYSIS PROCEDURES ................................................... 63 
5.1 GENERAL.................................................................................................................. 63 
5.2 EQUIVALENT LATERAL FORCE PROCEDURE ................................................. 65 
5.2.1 Seismic base shear ...................................................................................... 66 
5.2.2 Period determination.................................................................................. 68 
5.2.3 Vertical distribution of seismic forces ........................................................ 71 
5.2.4 Horizontal shear distribution....................................................................... 71 
5.2.5 Overturning................................................................................................. 74 
5.2.6 Drift determination and P-delta effects....................................................... 74 
5.3 RESPONSE SPECTRUM PROCEDURE.................................................................. 76 
5.3.2 Modes.......................................................................................................... 77 
5.3.3 Modal properties ......................................................................................... 77 
5.3.4 Modal base shear ........................................................................................ 77 
5.3.5 Modal forces, deflections and drifts............................................................ 78 
5.3.6 Modal story shears and moments................................................................ 78 
5.3.7 Design values.............................................................................................. 78 
5.3.8 Horizontal shear distribution....................................................................... 79 
5.3.9 Foundation overturning............................................................................... 79 
5.3.10 P-delta effects ........................................................................................... 79 
5.4 LINEAR RESPONSE HISTORY PROCEDURE...................................................... 79 
5.5 NONLINEAR RESPONSE HISTORY PROCEDURE ............................................. 80 
5.5.4 Design review ............................................................................................. 81 
5.6 SOIL-STRUCTURE INTERATION EFFECTS ....................................................... 82 
5.6.1 General........................................................................................................ 82 
5.6.2 Equivalent lateral force procedure ............................................................. 87 
5.6.3 Response spectrum procedure..................................................................... 96 
APPENDIX to Chapter 5, NONLINEAR STATIC PROCEDURE .................................. 103 
Chapter 6, ARCHITECTURAL MECHANICAL AND ELECTRICAL 
COMPONENT DESIGN REQUIREMENTS ................................................................... 109 
6.1 GENERAL................................................................................................................ 109 
6.1.1 Scope......................................................................................................... 109 
6.2 GENERAL DESIGN REQUIREMENTS ................................................................ 109 
6.2.2 Component importance factor................................................................... 109 
6.2.3 Consequential damage .............................................................................. 110 
6.2.4 Flexibility.................................................................................................. 111 
6.2.5 Component force transfer ......................................................................... 111 
6.2.6 Seismic forces ........................................................................................... 112 
6.2.7 Seismic relative displacements ................................................................. 115 
 v 
6.2.8 Component anchorage...............................................................................116 
6.2.9 Construction documents............................................................................117 
6.3 ARCHITECTURAL COMPONENTS......................................................................118 
6.3.1 Forces and displacements..........................................................................118 
6.3.2 Exterior nonstructural wall elements and connections..............................119 
6.3.3 Out-of-plane bending ................................................................................119 
6.3.4 Suspended ceilings ....................................................................................120 
6.3.5 Access floors .............................................................................................120 
6.3.6 Partitions ...................................................................................................121 
6.3.7 Glass I curtain walls, glazed storefronts and glazed partitions ................121 
6.4 MECHANICAL AND ELECTRICAL COMPONENTS .........................................128 
6.4.1 Component period .....................................................................................130 
6.4.2 Mechanical components............................................................................130 
6.4.3 Electrical components ...............................................................................130 
6.4.4 Supports and attachments..........................................................................131 
6.4.5 Utility and service lines.............................................................................131 
6.4.6 HVAC ductwork .......................................................................................132 
6.4.7 Piping systems...........................................................................................132 
6.4.8 Boilers and pressure vessels ......................................................................133 
6.4.9 Elevators....................................................................................................133 
Appendix to Chapter 6, ALTERNATIVE PROVISIONS FOR THE DESIGN 
OF PIPING SYSTEMS .....................................................................................................137 
Chapter 7 FOUNDATION DEIGN REQUIREMENTS ....................................................139 
7.1 GENERAL ................................................................................................................139 
7.1.1 Scope .........................................................................................................139 
7.2 GENERAL DESIGN REQUIREMENTS.................................................................139 
7.2.2 Soil capacities ...........................................................................................139 
7.2.3 Foundation load-deformation characteristics ............................................139 
7.3 SEISMIC DESIGN CATEGORY B.........................................................................139 
7.4 SEISMIC DESIGN CATEGORY C.........................................................................139 
7.4.1 Investigation ..............................................................................................139 
7.4.2 Pole-type structures ...................................................................................153 
7.4.3 Foundation ties ..........................................................................................153 
7.4.4 Special pile requirements ..........................................................................154 
7.5 SEISMIC DESIGN CATEGORIES D, E, and F ......................................................155 
7.5.1 Investigation .............................................................................................155 
7.5.3 Foundation ties .........................................................................................158 
7.5.4 Special pile and grade beam requirements ................................................158 
Appendix to Chapter 7, GEOTECHNICAL ULTIMATE STRENGTH DESIGN 
OF FOUNDATIONS AND FOUNDATION LOAD-DEFORMATION 
MODELING.......................................................................................................................167 
Chapter 8 STEEL STRUCTURE DESIGN REQUIREMENTS........................................175 
8.1 GENERAL ................................................................................................................175 
8.1.2 References .................................................................................................175 
8.2 GENERAL DESIGN REQUIREMENTS.................................................................175 
8.2.1 Seismic Design Categories B and C..........................................................175 
8.2.2 Seismic Design categories D, E, and F .....................................................175 
8.4 COLD-FORM STEEL ..............................................................................................175 
8.4.2 Light-frame walls ......................................................................................175 
 vi 
8.4.4 Steel deck diaphragms .............................................................................. 176 
8.5 STEEL CABLES ...................................................................................................... 176 
8.6 RECOMMENDED PROVISIONS FOR BUCKLING-RESTRAINED 
BRACED FRAMES ......................................................................................... 176 
8.6.3 Commentary on buckling-restrained braced frames ................................. 176 
8.7 SPECIAL STEEL PLATE WALLS ........................................................................ 189 
8.7.3 Scope......................................................................................................... 189 
8.7.4 Webs ......................................................................................................... 190 
8.7.5 Connections of webs to boundary elements.............................................. 192 
8.7.6 Horizontal and vertical boundary elements (HBE and VBE) ................... 192 
Chapter 9 CONCRETE STRUCTURE DESIGN REQUIREMENTS .............................. 199 
9.1 GENERAL................................................................................................................ 199 
9.1.2 References................................................................................................. 199 
9.2 GENERAL DESIGN REQUIREMENTS ................................................................ 199 
9.2.1 Classification of shear walls ..................................................................... 199 
9.2.2 Modifications to ACI 318 ......................................................................... 199 
9.3 SEISMIC DESIGN CATEGORY B......................................................................... 200 
9.3.1 Ordinary moment frames .......................................................................... 200 
9.4 SEISMIC DESIGN CATEGORY C......................................................................... 200 
9.5 SEISMIC DESIGN CATEGORIESD, E, AND F .................................................... 201 
9.6 ACCEPTANCE CRITERIA FOR SPECIAL PRECAST STRUCTURAL 
WALLS BASED ON VALIDATION TESTING ............................................ 200 
9.6.1 Notation .................................................................................................... 200 
9.6.2 Definitions ................................................................................................ 200 
9.6.3 Scope and general requirements ............................................................... 204 
9.6.4 Design procedure ...................................................................................... 206 
9.6.5 Test modules ............................................................................................. 207 
9.6.6 Testing agency .......................................................................................... 208 
9.6.7 Test method............................................................................................... 209 
9.6.8 Test report ................................................................................................. 211 
9.6.9 Test module acceptance criteria................................................................ 211 
Appendix to Chapter 9, UNTOPPED PRECAST DIPHRAGMS ..................................... 217 
Chapter 10, COMPOSITE STEEL AND CONCRETE STRUCTURE DESIGN 
REQUIREMENTS......................................................................................................... 221 
10.1 GENERAL ................................................................................................................ 221 
10.4 SEISMIC DESIGN CATEGORIES D, E, AND F .................................................... 221 
Chapter 11 MASONRY STRUCTURE DESIGN REQUIREMENTS.............................. 223 
11.1.2 References .............................................................................................. 223 
11.4 GLASS-UNIT MASONRY AND MASONRY VENEER..................................... 224 
11.5 PRESTRESSED MASONRY................................................................................. 224 
11.6 ANCHORING TO MASONRY ............................................................................. 224 
Chapter 12 WOOD STRUCTURE DESIGN REQUIREMENTS..................................... 231 
12.1 GENERAL.............................................................................................................. 231 
12.1.2 References............................................................................................... 231 
12.2 DESIGN METHODS.............................................................................................. 231 
12.2.1 Seismic design categories B, C, and D ................................................... 232 
 vii 
12.2.2 Seismic design Categories E and F .........................................................233 
12.3 GENERAL DESIGN REQUIREMENTS FOR ENGINEERED WOOD 
CONSTRUCTION............................................................................................238 
12.3.1 Framing ...................................................................................................255 
12.4 CONVENTIONAL LIGHT-FRAME CONSTRUCTION......................................255 
12.4.1 Limitations ..............................................................................................256 
12.4.2 Braced walls ............................................................................................257 
12.4.3 Detailing requirements ............................................................................258 
Chapter 13 SEISMICALLY ISOLATED STRUCTURE DESIGN 
REQUIREMENTS............................................................................................273 
13.1 GENERAL .............................................................................................................273 
13.1.1 Scope .......................................................................................................274 
13.2 GENERAL DESIGN REQUIREMENTS...............................................................275 
13.2.1 Occupancy importance factor..................................................................275 
13.2.4 Procedure selection .................................................................................275 
13.2.5 Isolation system.......................................................................................277 
13.2.6 Structural system.....................................................................................278 
13.3 EQUIVALENT LATERAL FORCE PROCEDURE..............................................278 
13.3.2 Minimum lateral displacements ..............................................................278 
13.3.3 Minimum distribution of forces ..............................................................280 
13.3.4 Vertical distribution of forces .................................................................280 
13.3.5 Drift limits ...............................................................................................281 
13.4 DYNAMIC PROCEDURES...................................................................................281 
13.5 DESIGN REVIEW..................................................................................................282 
13.6 TESTING................................................................................................................282 
13.6.4 Design properties of the isolation system................................................283 
Chapter 14 NONBUILDING STRURCTURE DESIGN REQUIREMENTS ...................287 
14.1 GENERAL ..............................................................................................................287 
14.1.1 Scope .......................................................................................................287 
14.1.2 References ...............................................................................................289 
14.1.5 Nonbuilding structures supported by other structures.............................289 
14.2 GENERAL DESIGN REQUIREMENTS...............................................................290 
14.2.1 Seismic Use groups and importance factors............................................290 
14.2.3 Design basis.............................................................................................293 
14.2.4 Seismic-force-resisting system selection and limitations........................293 
14.2.5 Structural analysis procedure selection ..................................................294 
14.2.9 Fundamental period.................................................................................294 
14.3 NONBUILDING STRUCTUTRES SIMILAR TO BUILDINGS ........................294 
14.3.1 Electrical power generating facilities ......................................................294 
14.3.3 Piers and wharves....................................................................................294 
14.3.4 Pipe racks ................................................................................................296 
14.3.5 Steel storage tanks...................................................................................296 
14.4 NONBUILDING STRUCTURES NOT SIMILAR TO BUILDINGS ...................296 
14.4.2 Earth retaining structures ........................................................................296 
14.4.3 Stacks and chimneys ...............................................................................296 
14.4.7 Tanks and vessels ....................................................................................297 
Appendix to Chapter 14 OTHER NONBUILDING STRUCTURES................................305 
Chapter 15 STRUCTURES WITH DAMPING SYSTEMS..............................................309 

 viii 
Commentary Appendix A, DEVELOPMENT OF MAXIMUM CONSIDERED 
EARTHQUAKE GROUND MOTION MAPS FIGURES 3.3-1 
THROUGH 3.3-14............................................................................................ 317 
Commentary Appendix B, DEVELOPMENT OF THE USGS SEISMIC 
MAPS ................................................................................................................ 331 
THE COUNCIL: ITS PURPOSE AND ACTIVITIES...................................................... 367 
BSSC MEMBER ORGANIZATIONS.............................................................................. 383 
BUILDING SEISMIC SAFETY COUNCIL PUBLICATIONS ....................................... 384 

 1 
Chapter 1 Commentary 
GENERAL PROVISIONS 
Chapter 1 sets forth general requirements for applying the analysis and design provisions contained in 
Chapters 2 through 14 of the NEHRP Recommended Provisions for Seismic Regulations for New 
Buildings and Other Structures. It is similar to what might be incorporated in a code as administrative 
regulations. 
Chapter 1 is designed to be as compatible as possible with normal code administrative provisions, but it 
is written as the guide to use of the rest of the document, not as a regulatory mechanism. The word 
shall is used in the Provisions not as a legal imperative, but simply as the language necessary to 
ensure fulfillment of all the steps necessary to technically meet a minimum standard of performance. 
It is important to note that the Provisions is intended to serve as a resource document for use by any 
interested member of the building community. Thus, some users may alter certain information within 
the Provisions (e.g., the determination of which use groups are included within the higher Seismic Use 
Groups might depend on whether the user concluded that the generally more-demanding design 
requirements were necessary). It is strongly emphasized, however, that such tailoring should be 
carefully considered by highly qualified individuals who are fully aware of all the implications of any 
changes on all affected procedures in the analysis and design sequences of the document. 
Further, although the Provisions is national in scope, it presents minimum criteria. It is neither intended 
to nor does it justify any reduction in higher standards that have been locally established, particularly in 
areas of highest seismicity. 
Reference is made throughout the document to decisions and actions that are delegated to an unspecified 
authority having jurisdiction. The document is intended to be applicable to many different types of 
jurisdictions and chains of authority, and an attempt has been made to recognize situations where more 
than technical decision-making can be presumed. In fact, the document anticipates the need to establish 
standards and approval systems to accommodate the use of the document for development of a 
regulatory system. A good example of this is in Sec. 1.1.2.5 where the need for well-established criteria 
and systems of testing and approval are recognized even though few such systems are in place. In some 
instances, the decision-making mechanism referred to is clearly most logically the province of a 
building official or department; in others, it may be a law-making body such as a state legislature, a city 
council, or some other state or local policy-making body. The term authority having jurisdiction has 
been used to apply to all of these entities. A good example of the need for keeping such generality in 
mind is provided by the California law concerning the design and construction of schools. That law 
establishes requirements for independent special inspection approved and supervised by the Office of 
the State Architect, a state-level office that does not exist in many other states. 
Note that Appendix A to this Commentary volume presents a detailed explanation of the development of 
Provisions Maps 1 through 24 and Appendix B describes development of the U.S. Geological Survey 
seismic hazard maps on which the Provisions maps are based. An overview of the Building Seismic 
Safety Council (BSSC) and its activities appears at the end of the volume. 
1.1 GENERAL 
1.1.1 Purpose. The goal of the Provisions is to present criteria for the design and construction of new 
structures subject to earthquake ground motions in order to minimize the hazard to life for all structures, 
to increase the expected performance of structures having a substantial public hazard due to occupancy 
or use as compared to ordinary structures, and to improve the capability of essential facilities to function 

2003 Commentary, Chapter 1 
2 
after an earthquake. To this end, the Provisions provides the minimum criteria considered prudent for 
the protection of life safety in structures subject to earthquakes. The Provisions document has been 
reviewed extensively and balloted by the architectural, engineering, and construction communities and, 
therefore, it is a proper source for the development of building codes in areas of seismic exposure. 
Some design standards go further than the Provisions and attempt to minimize damage as well as protect 
building occupants. For example, the California Building Code has added property protection in 
relation to the design and construction of hospitals and public schools. The Provisions document 
generally considers property damage as it relates to occupant safety for ordinary structures. For high 
occupancy and essential facilities, damage limitation criteria are more strict in order to better provide for 
the safety of occupants and the continued functioning of the facility. 
Some structural and nonstructural damage can be expected as a result of the design ground motions 
because the Provisions allow inelastic energy dissipation in the structural system. For ground motions 
in excess of the design levels, the intent of the Provisions is for the structure to have a low likelihood of 
collapse. 
It must be emphasized that absolute safety and no damage even in an earthquake event with a 
reasonable probability of occurrence cannot be achieved for most structures. However, a high degree of 
life safety, albeit with some structural and nonstructural damage, can be achieved economically in 
structures by allowing inelastic energy dissipation in the structure. The objective of the Provisions 
therefore is to set forth the minimum requirements to provide reasonable and prudent life safety. For 
most structures designed and constructed according to the Provisions, it is expected that structural 
damage from even a major earthquake would likely be repairable, but the damage may not be 
economically repairable. 
Where damage control is desired, the design must provide not only sufficient strength to resist the 
specified seismic loads but also the proper stiffness to limit the lateral deflection. Damage to 
nonstructural elements may be minimized by proper limitation of deformations; by careful attention to 
detail; and by providing proper clearances for exterior cladding, glazing, partitions, and wall panels. 
The nonstructural elements can be separated or floated free and allowed to move independently of the 
structure. If these elements are tied rigidly to the structure, they should be protected from deformations 
that can cause cracking; otherwise, one must expect such damage. It should be recognized, however, 
that major earthquake ground motions can cause deformations much larger than the specified drift limits 
in the Provisions. 
Where prescribed wind loading governs the stress or drift design, the resisting system still must conform 
to the special requirements for seismic-force-resisting systems. This is required in order to resist, in a 
ductile manner, potential seismic loadings in excess of the prescribed loads. 
A proper, continuous load path is an obvious design requirement for equilibrium, but experience has 
shown that it often is overlooked and that significant damage and collapse can result. The basis for this 
design requirement is twofold: 
1. To ensure that the design has fully identified the seismic-force-resisting system and its appropriate 
design level and 
2. To ensure that the design basis is fully identified for the purpose of future modifications or changes 
in the structure. 
Detailed requirements for selecting or identifying and designing this load path are given in the 
appropriate design and materials chapters. 
1.1.2.1 Scope. The scope statement establishes in general terms the applicability of the Provisions as a 
base of reference. Certain structures are exempt and need not comply: 
General Provisions 
3 
1. Detached one- and two-family dwellings in Seismic Design Categories A, B, and C are exempt 
because they represent low seismic risks. 
2. Structures constructed using the conventional light-frame construction requirements in Sec. 12.5 are 
deemed capable of resisting the seismic forces imposed by the Provisions. While specific elements 
of conventional light-frame construction may be calculated to be overstressed, there is typically a 
great deal of redundancy and uncounted resistance in such structures. Detached one- and two-story 
wood-frame dwellings have generally performed well even in regions of higher seismicity. The 
requirements of Sec. 12.5 are adequate to provide the safety required for such dwellings without 
imposing any additional requirements of the Provisions. 
3. Agricultural storage structures are generally exempt from most code requirements because of the 
exceptionally low risk to life involved and that is the case of the Provisions. 
4. Structures in areas with extremely low seismic risk need only comply with the design and detailing 
requirements for structures assigned to Seismic Design Category A. 
The Provisions are not retroactive and apply only to existing structures when there is an addition, 
change of use, or alteration. As a minimum, existing structures should comply with legally adopted 
regulations for repair and rehabilitation as related to earthquake resistance. (Note: Publications such as 
the Handbook for the Seismic Evaluation of BuildingsA Prestandard [FEMA 310] and the 
Prestandard and Commentary for the Seismic Rehabilitation of Buildings [FEMA 356] are available.) 
The Provisions are not written to prevent damage due to earth slides (such as those that occurred in 
Anchorage, Alaska), to liquefaction (such as occurred in Niigata, Japan), or to tsunami (such as 
occurred in Hilo, Hawaii). It provides for only minimum required resistance to earthquake ground 
shaking, without settlement, slides, subsidence, or faulting in the immediate vicinity of the structure. 
1.1.2.2 Additions. Additions that are structurally independent of an existing structure are considered to 
be new structures required to comply with the Provisions. For additions that are not structurally 
independent, the intent is that the addition as well as the existing structure be made to comply with the 
Provisions except that an increase of up to 5 percent of the mass contributing to seismic forces is 
permitted in any elements of the existing structure without bringing the entire structure into compliance 
with the Provisions. Additions also shall not reduce the lateral force resistance of any existing element 
to less than that required for a new structure. 
1.1.2.3 Change of use. When a change in the use of a structure will result in the structure being 
reclassified to a higher Seismic Use Group, the existing structure must be brought into compliance with 
the requirements of the Provisions as if it were a new structure. Structures in higher Seismic Use 
Groups are intended to provide a higher level of safety to occupants and in the case of Seismic Use 
Group III to be capable of performing their safety-related function after a seismic event. An exception 
is allowed when the change is from Seismic Use Group I to Seismic Use Group II where SDS is less than 
0.3. The expense that may be necessary to upgrade such a structure because of a change in the Seismic 
Use Group cannot be justified for structures located in regions with low seismic risk. 
1.1.2.4 Alterations. Alterations include all significant modifications to existing structures that are not 
classified as an addition. No reduction in strength of the seismic-force-resisting system or stiffness of 
the structure shall result from an alteration unless the altered structure is determined to be in compliance 
with the Provisions. 
Like additions, an increase of not greater than 5 percent of the mass contributing to seismic forces is 
permitted in any structural element of the existing structure without bringing the entire structure into 
compliance with the Provisions. 
2003 Commentary, Chapter 1 
4 
The cumulative effects of alterations and additions should not increase the seismic forces in any 
structural element of the existing structure by more than 5 percent unless the capacity of the element 
subject to the increased seismic forces is still in compliance with the Provisions. 
1.1.2.5 Alternate materials and alternate means and methods of construction. It is not possible for 
a design standard to provide criteria for the use of all possible materials and their combinations and 
methods of construction either existing or anticipated. While not citing specific materials or methods of 
construction currently available that require approval, this section serves to emphasize the fact that the 
evaluation and approval of alternate materials and methods require a recognized and accepted approval 
system. The requirements for materials and methods of construction contained within the document 
represent the judgment of the best use of the materials and methods based on well-established expertise 
and historical seismic performance. It is important that any replacement or substitute be evaluated with 
an understanding of all the ramifications of performance, strength, and durability implied by the 
Provisions. 
It also is recognized that until needed approval standards and agencies are created, authorities having 
jurisdiction will have to operate on the basis of the best evidence available to substantiate any 
application for alternates. If accepted standards are lacking, it is strongly recommended that 
applications be supported by extensive reliable data obtained from tests simulating, as closely as is 
practically feasible, the actual load and/or deformation conditions to which the material is expected to 
be subjected during the service life of the structure. These conditions, where applicable, should include 
several cycles of full reversals of loads and deformations in the inelastic range. 
1.2 SEISMIC USE GROUPS 
The expected performance of structures shall be controlled by assignment of each structure to one of 
three Seismic Use Groups. Seismic Use Groups are categorized based on the occupancy of the 
structures within the group and the relative consequences of earthquake-induced damage to the 
structures. The Provisions specify progressively more conservative strength, drift control, system 
selection, and detailing requirements for structures contained in the three groups, in order to attain 
minimum levels of earthquake performance suitable to the individual occupancies. 
In previous editions of the Provisions, this categorization of structures, by occupancy, or use, was 
termed a Seismic Hazard Exposure Group. The name Seismic Use Group was adopted in the 1997 
Provisions as being more representative of the definition of this classification. Seismic hazard relates to 
the severity and frequency of ground motion expected to affect a structure. Since structures contained 
in these groups are spread across the various zones of seismicity, from high to low hazard, the groups do 
not really relate to hazard. Rather the groups, categorized by occupancy or use, are used to establish 
design criteria intended to produce specific types of performance in design earthquake events, based on 
the importance of reducing structural damage and improving life safety. 
In terms of post-earthquake recovery and redevelopment, certain types of occupancies are vital to public 
needs. These special occupancies were identified and given specific recognition. In terms of disaster 
preparedness, regional communication centers identified as critical emergency services should be in a 
higher classification than retail stores, office buildings, and factories. 
Specific consideration is given to Group III, essential facilities required for post-earthquake recovery. 
Also included are structures that contain substances, that if released into the environment, are deemed to 
be hazardous to the public. The 1991 Edition included a flag to urge consideration of the need for utility 
services after an earthquake. It is at the discretion of the authority having jurisdiction which structures 
are required for post-earthquake response and recovery. This is emphasized with the term designated 
before many of the structures listed in Sec. 1.2.1. Using Item 3, designated medical facilities having 
emergency treatment facilities, as an example, the authority having jurisdiction should inventory 
medical facilities having emergency treatment facilities within the jurisdiction and designate those to be 
General Provisions 
5 
required for post-earthquake response and recovery. In a rural location where there may not be a major 
hospital, the authority having jurisdiction may choose to require outpatient surgery clinics to be 
designated Group III structures. On the other hand, these same clinics in a major jurisdiction with 
hospitals nearby may not need to be designated Group III structures. 
Group II structures are those having a large number of occupants and those where the occupants ability 
to exit is restrained. The potential density of public assembly uses in terms of number of people warrant 
an extra level of care. The level of protection warranted for schools, day care centers, and medical 
facilities is greater than the level of protection warranted for occupancies where individuals are 
relatively self-sufficient in responding to an emergency. 
Group I contains all uses other than those excepted generally from the requirements in Sec. 1.1.2.1. 
Those in Group I have lesser life hazard only insofar as there is the probability of fewer occupants in the 
structures and the structures are lower and/or smaller. 
In structures with multiple uses, the 1988 Edition of the Provisions required that the structure be 
assigned the classification of the highest group occupying 15 percent or more of the total area of the 
structure. This was changed in the 1991 Edition to require the structure to be assigned to the highest 
group present. These requirements were further modified to allow different portions of a structure to be 
assigned different Seismic Use Groups provided the higher group is not negatively impacted by the 
lower group. When a lower group impacts a higher group, the higher group must either be seismically 
independent of the other, or the two must be in one structure designed seismically to the standards of the 
higher group. Care must be taken, however, for the case in which the two uses are seismically 
independent but are functionally dependent. The fire and life-safety requirements relating to exiting, 
occupancy, fire-resistive construction and the like of the higher group must not be reduced by 
interconnection to the lower group. Conversely, one must also be aware that there are instances, 
although uncommon, where certain fire and life-safety requirements for a lower group may be more 
restrictive than those for the higher group. Such assignments also must be considered when changes are 
made in the use of a structure even though existing structures are not generally within the scope of the 
Provisions. 
Consideration has been given to reducing the number of groupings by combining Groups I and II and 
leaving Group III the same as is stated above; however, the consensus of those involved in the 
Provisions development and update efforts to date is that such a merging would not be responsive to the 
relative performance desired of structures in these individual groups. 
Although the Provisions explicitly require design for only a single level of ground motion, it is expected 
that structures designed and constructed in accordance with these requirements will generally be able to 
meet a number of performance criteria, when subjected to earthquake ground motions of differing 
severity. The performance criteria discussed here were jointly developed during the BSSC Guidelines 
and Commentary for Seismic Rehabilitation of Buildings Project (ATC, 1995) and the Structural 
Engineers Association of California Vision 2000 Project (SEAOC, 1995). In the system established by 
these projects, earthquake performance of structures is defined in terms of several standardized 
performance levels and reference ground motion levels. Each performance level is defined by a limiting 
state in which specified levels of degradation and damage have occurred to the structural and 
nonstructural building components. The ground motion levels are defined in terms of their probability 
of exceedance. 
Although other terminology has been used in some documents, four performance levels are commonly 
described as meaningful for the design of structures. These may respectively be termed the operational, 
immediate occupancy, life safety, and collapse prevention levels. Of these, the operational level 
represents the least level of damage to the structure. Structures meeting this level when responding to 
an earthquake are expected to experience only negligible damage to their structural systems and minor 
damage to nonstructural systems. The structure will retain nearly all of its pre-earthquake strength and 
2003 Commentary, Chapter 1 
6 
stiffness and all mechanical, electrical, plumbing, and other systems necessary for the normal operation 
of the structure are expected to be functional. If repairs are required, these can be conducted at the 
convenience of the occupants. 
The risk to life safety during an earthquake in a structure meeting this performance level is negligible. 
Note, that in order for a structure to meet this level, all utilities required for normal operation must be 
available, either through standard public service or emergency sources maintained for that purpose. 
Except for very low levels of ground motion, it is generally not practical to design structures to meet 
this performance level. 
The immediate occupancy level is similar to the operational level although somewhat more damage to 
nonstructural systems is anticipated. Damage to the structural systems is very slight and the structure 
retains all of its pre-earthquake strength and nearly all of its stiffness. Nonstructural elements, including 
ceilings, cladding, and mechanical and electrical components, remain secured and do not represent 
hazards. Exterior nonstructural wall elements and roof elements continue to provide a weather barrier, 
and to be otherwise serviceable. The structure remains safe to occupy; however, some repair and 
clean-up is probably required before the structure can be restored to normal service. In particular, it is 
expected that utilities necessary for normal function of all systems will not be available, although those 
necessary for life safety systems would be provided. Some equipment and systems used in normal 
function of the structure may experience internal damage due to shaking of the structure, but most 
would be expected to operate if the necessary utility service was available. Similar to the operational 
level, the risk to life safety during an earthquake in a structure meeting this performance level is 
negligible. Structural repair may be completed at the occupants convenience, however, significant 
nonstructural repair and cleanup is probably required before normal function of the structure can be 
restored. 
At the life safety level, significant structural and nonstructural damage has occurred. The structure may 
have lost a substantial amount of its original lateral stiffness and strength but still retains a significant 
margin against collapse. The structure may have permanent lateral offset and some elements of the 
seismic-force-resisting system may exhibit substantial cracking, spalling, yielding, and buckling. 
Nonstructural elements of the structure, while secured and not presenting falling hazards, are severely 
damaged and cannot function. The structure is not safe for continued occupancy until repairs are 
instituted as strong ground motion from aftershocks could result in life threatening damage. Repair of 
the structure is expected to be feasible, however, it may not be economically attractive to do so. The 
risk to life during an earthquake, in a structure meeting this performance level is very low. 
At the collapse prevention level a structure has sustained nearly complete damage. The seismic-forceresisting 
system has lost most of its original stiffness and strength and little margin remains against 
collapse. Substantial degradation of the structural elements has occurred including extensive cracking 
and spalling of masonry and concrete elements and buckling and fracture of steel elements. The 
structure may have significant permanent lateral offset. Nonstructural elements of the structure have 
experienced substantial damage and may have become dislodged creating falling hazards. The structure 
is unsafe for occupancy as even relatively moderate ground motion from aftershocks could induce 
collapse. Repair of the structure and restoration to service is probably not practically achievable. 
The design ground motion contained in the Provisions is taken as two-thirds of the maximum 
considered earthquake ground motion. Such ground motion may have a return period varying from a 
few hundred years to a few thousand years, depending on the regional seismicity. It is expected that 
structures designed in accordance with the requirements for Group I would achieve the life safety or 
better performance level for these ground motions. Structures designed in accordance with the 
requirements for Group III should be able to achieve the Immediate Occupancy or better performance 
level for this ground motion. Structures designed to the requirements for Group II would be expected to 
achieve performance better than the life safety level but perhaps less than the immediate occupancy 
level for this ground motion. 
General Provisions 
7 
While the design ground motion represents a rare earthquake event, it may not be the most severe event 
that could ever affect a site. In zones of moderate seismicity, it has been common practice in the past to 
consider ground motion with a 98 percent chance of non-exceedance in 50 years, or an average return 
period of 2,500 years, as being reasonably representative of the most severe ground motion ever likely 
to affect a site. This earthquake has been variously termed a maximum credible earthquake, maximum 
capable event and, most recently, a maximum considered earthquake. The recent terminology is 
adopted here in recognition that ground motion of this probability level is not the most severe motion 
that could ever effect the site, but is considered sufficiently improbable that more severe ground motions 
need not practically be considered. In regions near major active faults, such as coastal California, 
estimates of ground motion at this probability of exceedance can produce structural demands much 
larger than has typically been recorded in past earthquakes. Consequently, in these zones, the maximum 
considered earthquake is now commonly taken based on conservative estimates of the ground motion 
from a deterministic event, representing the largest magnitude event that the nearby faults are believed 
capable of producing. 
It is expected that structures designed to the requirements for Group I would be capable of responding to 
the maximum considered earthquake at a near collapse or better performance level. Structures designed 
to the requirements for Group III should be capable of responding to such ground motions at the life 
safety level. Structures designed and constructed to the requirements for Group II structures should be 
capable of responding to maximum considered earthquake ground motions with a performance 
intermediate to the near collapse and life safety levels. 
In zones of high seismicity, structures may experience strong motion earthquakes several times during 
their lives. It is also important to consider the performance expected of structures for these somewhat 
less severe, but much more frequent, events. For this purpose, earthquake ground shaking with a 50 
percent probability of non-exceedance in 50 years may be considered. Sometimes termed a maximum 
probable event (MPE), such ground motion would be expected to recur at a site, one time, every 72 
years. Structures designed to the requirements for Group I would be expected to respond to such ground 
motion at the Immediate Occupancy level. Structures designed and constructed to either the Group II or 
Group III requirements would be expected to perform to the Operational level for these events. This 
performance is summarized in Figure C1.2-1. 
It is important to note that while the performance indicated in Figure C1.2-1 is generally indicative of 
that expected for structures designed in accordance with the Provisions, there can be significant 
variation in the performance of individual structures from these expectations. This variation results 
Figure 1.2-1 Expected building performance. 
2003 Commentary, Chapter 1 
8 
from individual site conditions, quality of construction, structural systems, detailing, overall 
configuration of the structure, inaccuracies in our analytical techniques, and a number of other complex 
factors. As a result of these many factors, and intentional conservatism contained in the Provisions, 
most structures will perform better than indicated in the figure and others will not perform as well. 
1.2.5 Seismic Use Group III structure access protection. This section establishes the requirement for 
access protection for Seismic Use Group III structures. There is a need for ingress/egress to those 
structures that are essential post-earthquake facilities and this shall be considered in the siting and 
design of the structure. 
1.3 OCCUPANCY IMPORTANCE FACTOR 
Although the concept of an occupancy importance factor for structural systems has been included in the 
Uniform Building Code for many years, it was first adopted into the 1997 Edition of the Provisions. 
The inclusion of the occupancy importance factor is one of several requirements included in this edition 
of the Provisions where there are attempts to control the seismic performance capability of structures in 
the different Seismic Use Groups. Specifically, the occupancy importance factor modifies the R 
coefficients used to determine minimum design base shears. Structures assigned occupancy importance 
factors greater than 1.0 must be designed for larger seismic forces. As a result, these structures are 
expected to experience lower ductility demands than structures designed with lower occupancy 
importance factors and, thus sustain less damage. The Provisions also include requirements that attempt 
to limit vulnerability to structural damage by specifying more stringent drift limits for structures in 
Seismic Use Groups of higher risk. Further discussion of these concepts is found in Commentary 
Sec. 4.2.1 and 4.5. 
1.4 SEISMIC DESIGN CATEGORY 
This section establishes the design categories that are the keys for establishing design requirements for 
any structure based on its use (Seismic Use Group) and on the level of expected seismic ground motion. 
Once the Seismic Design Category (A, B, C, D, E, or F) for the structure is established, many other 
requirements such as detailing, quality assurance, system limits, height limitations, specialized 
requirements, and change of use are related to it. 
Prior to the 1997 edition of the Provisions, these categories were termed Seismic Performance 
Categories. While the desired performance of the structure, under the design earthquake, was one 
consideration used to determine which category a structure should be assigned to, it was not the only 
factor. The seismic hazard at the site was actually the principle parameter that affected a structures 
category. The name was changed to Seismic Design Category to represent the uses of these categories, 
which is to determine the specific design requirements. 
The earlier editions of the Provisions utilized the peak velocity-related acceleration, Av, to determine a 
buildings Seismic Performance Category. However, this coefficient does not adequately represent the 
damage potential of earthquakes on sites with soil conditions other than rock. Consequently, the 1997 
Provisions adopted the use of response spectral acceleration parameters SDS and SD1, which include site 
soil effects for this purpose. Instead of a single table, as was present in previous editions of the 
Provisions, two tables are now provided, relating respectively to short-period and long-period ground 
motions. 
Seismic Design Category A represents structures in regions where anticipated ground motions are 
minor, even for very long return periods. For such structures, the Provisions require only that a 
complete seismic-force-resisting system be provided and that all elements of the structure be tied 
together. A nominal design force equal to 1 percent of the weight of the structure is used to proportion 
the lateral system. 
It is not considered necessary to specify seismic-resistant design on the basis of a maximum considered 
earthquake ground motion for Seismic Design Category A structures because the ground motion 
General Provisions 
9 
computed for the areas where these structures are located is determined more by the rarity of the event 
with respect to the chosen level of probability than by the level of motion that would occur if a small but 
close earthquake actually did occur. However, it is desirable to provide some protection against 
earthquakes and many other types of unanticipated loadings. Thus, the requirements for Seismic Design 
Category A provide a nominal amount of structural integrity that will improve the performance of 
buildings in the event of a possible but rare earthquake even though it is possible that the ground 
motions could be large enough to cause serious damage or even collapse. The result of design to 
Seismic Design Category A requirements is that fewer building would collapse in the vicinity of such an 
earthquake. 
The integrity is provided by a combination of requirements. First, a complete load path for lateral 
forces must be identified. Then it must be designed for a lateral force based on a 1 percent acceleration 
of the mass. The minimum connection forces specified for Seismic Design Category A also must be 
satisfied. 
The 1 percent value has been used in other countries as a minimum value for structural integrity. For 
many structures, design for the wind loadings specified in the local buildings codes normally will 
control the lateral force design when compared to the minimum integrity force on the structure. 
However, many low-rise, heavy structures or structures with significant dead loads resulting from heavy 
equipment may be controlled by the nominal 1 percent acceleration. Also, minimum connection forces 
may exceed structural forces due to wind in some structures. 
Seismic Design Category B includes Seismic Use Group I and II structures is regions of seismicity 
where only moderately destructive ground shaking is anticipated. In addition to the requirements for 
Seismic Design Category A, structures in Seismic Design Category B must be designed for forces 
determined using Maps 1 through 24. 
Seismic Design Category C includes Seismic Use Group III structures in regions where moderately 
destructive ground shaking may occur as well as Seismic Use Group I and II structures in regions with 
somewhat more severe ground shaking potential. In Seismic Design Category C, the use of some 
structural systems is limited and some nonstructural components must be specifically designed for 
seismic resistance. 
Seismic Design Category D includes structures of Seismic Use Group I, II, and III located in regions 
expected to experience destructive ground shaking but not located very near major active faults. In 
Seismic Design Category D, severe limits are placed on the use of some structural systems and irregular 
structures must be subjected to dynamic analysis techniques as part of the design process. 
Seismic Design Category E includes Seismic Use Group I and II structures in regions located very close 
to major active faults and Seismic Design Category F includes Seismic Use Group III structures in these 
locations. Very severe limitations on systems, irregularities, and design methods are specified for 
Seismic Design Categories E and F. For the purpose of determining if a structure is located in a region 
that is very close to a major active fault, the Provisions use a trigger of a mapped maximum considered 
earthquake spectral response acceleration parameter at 1-second period, S1, of 0.75 or more regardless of 
the structures fundamental period. The mapped short period acceleration, SS, was not used for this 
purpose because short period response accelerations do not tend to be affected by near-source conditions 
as strongly as do response accelerations at longer periods. 
Local or regional jurisdictions enforcing building regulations need to consider the effect of the maps, 
typical soil conditions, and Seismic Design Categories on the practices in their jurisdictional areas. For 
reasons of uniformity of practice or reduction of potential errors, adopting ordinances could stipulate 
particular values of ground motion, particular Site Classes, or particular Seismic Design Categories for 
all or part of the area of their jurisdiction. For example: 
2003 Commentary, Chapter 1 
10 
1. An area with an historical practice of high seismic zone detailing might mandate a minimum 
Seismic Design Category of D regardless of ground motion or Site Class. 
2. A jurisdiction with low variation in ground motion across the area might stipulate particular values 
of the ground motion rather than requiring use of the maps. 
3. An area with unusual soils might require use of a particular Site Class unless a geotechnical 
investigation proves a better Site Class. 
There are two limits on period for permission to ignore SD1 when establishing the Seismic Design 
Category. The first rule, requiring Ta be less than 80% of Ts, allows some conservatism for the 
uncertainty in estimating periods. The second rule only applies where a different period is used for 
computing drift than for computing forces. In that case, the period used for establishing drift must be 
less than the corner period, Ts. It should be noted that the period used for establishing drift could simply 
be Ta and, as such, does not require that the actual building period be calculated. 
1.4.2 Site limitation for Seismic Design Categories E and F. The forces that result on a structure 
located astride the trace of a fault rupture that propagates to the surface are extremely large and it is not 
possible to reliably design a structure to resist such forces. Consequently, the requirements of this 
section limit the construction of buildings in Seismic Design Categories E and F on sites subject to this 
hazard. Similarly, the effects of landsliding, liquefaction, and lateral spreading can be highly damaging 
to a building. However, the effects of these site phenomena can more readily be mitigated through the 
incorporation of appropriate design measures than can direct ground fault rupture. Consequently, 
construction on sites with these hazards is permitted if appropriate mitigation measures are included in 
the design. 
1.5 REQUIREMENTS FOR SEISMIC DESIGN CATEGORY A 
Because of the very low seismicity associated with sites with SDS less than 0.25 and SD1 less than 0.10, it 
is considered appropriate for Category A buildings to require only a complete seismic-force-resisting 
system, good quality of construction materials and adequate ties and anchorage as specified in this 
section. Category A buildings will be constructed in a large portion of the United States that is 
generally subject to strong winds but low earthquake risk. Those promulgating construction regulations 
for these areas may wish to consider many of the low-level seismic requirements as being suitable to 
reduce the windstorm risk. Since the Provisions considers only earthquakes, no other requirements are 
prescribed for Category A buildings. Only a complete seismic-force-resisting system, ties, and wall 
anchorage are required by these Provisions. 
Construction qualifying under Category A may be built with no special detailing requirements for 
earthquake resistance. Special details for ductility and toughness are not required in Category A. 
1.5.1 Lateral forces. This analysis procedure, which was added to the Provisions in the 1997 edition, 
is applicable only to structures in Seismic Design Category A. Such structures are not designed for 
resistance to any specific level of earthquake ground shaking as the probability that they would ever 
experience shaking of sufficient intensity to cause life threatening damage is very low so long as the 
structures are designed with basic levels of structural integrity. Minimum levels of structural integrity 
are achieved in a structure by assuring that all elements in the structure are tied together so that the 
structure can respond to shaking demands in an integral manner and also by providing the structure with 
a complete seismic-force-resisting system. It is believed that structures having this level of integrity 
would be able to resist, without collapse, the very infrequent earthquake ground shaking that could 
affect them. In addition, requirements to provide such integrity provides collateral benefit with regard 
to the ability of the structure to survive other hazards such as high wind storms, tornadoes, and 
hurricanes. 
The procedure outlined in this section is intended to be a simple approach to ensuring both that a 
building has a complete seismic-force-resisting system and that it is capable of sustaining at least a 
General Provisions 
11 
minimum level of lateral force. In this analysis procedure, a series of static lateral forces equal to 1 
percent of the weight at each level of the structure is applied to the structure independently in each of 
two orthogonal directions. The structural elements of the seismic-force-resisting system then are 
designed to resist the resulting forces in combination with other loads under the load combinations 
specified by the building code. 
The selection of 1 percent of the building weight as the design force for Seismic Design Category A 
structures is somewhat arbitrary. This level of design lateral force was chosen as being consistent with 
prudent requirements for lateral bracing of structures to prevent inadvertent buckling under gravity 
loads and also was believed to be sufficiently small as to not present an undue burden on the design of 
structures in zones of very low seismic activity. 
The seismic weight W is the total weight of the building and that part of the service load that might 
reasonably be expected to be attached to the building at the time of an earthquake. It includes 
permanent and movable partitions and permanent equipment such as mechanical and electrical 
equipment, piping, and ceilings. The normal human live load is taken to be negligibly small in its 
contribution to the seismic lateral forces. Buildings designed for storage or warehouse usage should 
have at least 25 percent of the design floor live load included in the weight, W. Snow loads up to 30 psf 
(1400 Pa) are not considered. Freshly fallen snow would have little effect on the lateral force in an 
earthquake; however, ice loading would be more or less firmly attached to the roof of the building and 
would contribute significantly to the inertia force. For this reason, the effective snow load is taken as 
the full snow load for those regions where the snow load exceeds 30 psf with the proviso that the local 
authority having jurisdiction may allow the snow load to be reduced up to 80 percent. The question of 
how much snow load should be included in W is really a question of how much ice buildup or snow 
entrapment can be expected for the roof configuration or site topography, and this is a question best left 
to the discretion of the local authority having jurisdiction. 
1.5.2 Connections. The requirements in this section are a simplified version of the material found in 
Sec. 4.6.1.1. For Seismic Design Category A, 5 percent is always greater than 0.133 times SDS. 
1.5.3 Anchorage of concrete or masonry walls. The intent of this section is to ensure that out-ofplane 
inertia forces generated within a concrete or masonry wall can be transferred to the adjacent roof 
or floor construction. The transfer can be accomplished only by reinforcement or anchors. 

2003 Commentary, Chapter 1 
12 
This page is intentionally left blank. 

13 
Chapter 2 Commentary 
QUALITY ASSURANCE 
2.1 GENERAL 
2.1.1 Scope. Quality assurance (control and verification) for structures assigned to Seismic Design 
Categories C, D, E and F, is necessary due to the complexity of the seismic-force-resisting systems and 
is important because of the serious consequences of the failure of structures. The level of quality 
assurance varies with the degree of seismic risk. 
Quality Assurance requirements involve many aspects of the structural design and construction 
processfrom the selection of the design team and their suitability for the project to the capabilities of 
the construction contractor(s) and subcontractors, whether selected by qualification or by low bid. 
Where structures are to be located in areas with a high probability of having damaging earthquake 
ground motion, adequate quality assurance is required to provide life safety. Unfortunately, in recent 
seismic events there have been numerous earthquake-related failures that are directly traceable to poor 
design or poor quality control during construction; these deficiencies must be eliminated. The 
earthquake requirements included in the Provisions rely heavily upon the concept of adequate quality 
control and verification to assure sound construction. It is important that all parties involved in the 
design and construction process understand and support the quality assurance requirements 
recommended in the Provisions. 
The technological complexity of the design of modern structures necessitates employment of a team of 
registered design professionals. Each member in responsible charge of design of each element or 
system of the structure must be qualified and licensed by the jurisdiction to practice in their technical 
fields of practice. Structures located at a site with a potential to have damaging earthquake ground 
motion must be designed to withstand the resulting seismic forces and accommodate element 
displacements. 
Every element of a structure is a part of a continuous load path transmitting seismic forces from and to 
the foundations, which must be adequately strengthened and appropriately anchored to resist the seismic 
forces and to accommodate the resulting displacements. Many of the failures in recent earthquakes have 
been attributed to weak links in the seismic-force-resisting load paths. Since the connections between 
adjacent elements of the structure often involve different registered design professionals and different 
construction trades during installation, it is imperative that these connections be adequately described in 
the construction documents and observed during installation. In order to accommodate these constraints 
and produce a coordinated design, the registered design professionals must function as an integrated and 
well coordinated team. 
The selection of the size and configuration of the structure, and the type of structural seismic-forceresisting 
system(s) selected, can have a significant impact on the performance of the structure in an 
earthquake. Since the selection can affect the design and cost of construction of almost every element 
of the structure, it is essential that the entire design team participate in making these preliminary design 
decisions and appropriately accommodate them in their design. While not required by the Provisions, it 
is recommended that a quality assurance plan be prepared for the design process. 
For quality assurance during construction, the following is included in the Provisions: (1) the registered 
design professional(s) in responsible charge of the design specifies the quality assurance requirements; 
(2) the prime contractor(s) exercises the control necessary to achieve the required quality; and (3) the 
owner monitors the construction process by means of consultants who perform special inspections, 
observations, and testing. It is important that all of the parties involved recognize their responsibilities, 
understand the procedures, and are capable of carrying them out. Because the contractor and specialty 
subcontractors are performing the work and exercising control of quality, it is essential that the special 
inspections and tests be performed by someone not in their direct employ. For this reason, the special 
2 003 Commentary, Chapter 2 
14 
inspectors are the owners inspectors and serve at the discretion of the authority having jurisdiction. 
When the owner is also the contractor, the owner, to avoid a potential conflict of interest, must engage 
independent agencies to conduct the special inspections and tests rather than try to qualify his own 
employees for that purpose. 
The contractual responsibilities during the construction phase vary from project to project depending on 
the structure, and the desires of the owner. The majority of building owners use the standard contract 
forms published by the American Institute of Architects (AIA) or the Engineers Joint Contract 
Documents Committee (EJCDC) (or a contract modeled therefrom) which include specific construction 
phase responsibilities. 
The registered design professional in responsible charge for each portion of the project is the most 
knowledgeable person available for assuring appropriate conformance with the intent of the design as 
conveyed in the construction documents. It is essential that a registered design professional be 
sufficiently involved during the construction phase of the project to assure general conformance with 
the approved construction documents. Courts are ruling more frequently that the above responsibilities 
remain that of the registered design professional in responsible charge of the design regardless of the 
language included in the contract for professional services. 
The quality assurance requirements included in Chapter 2 of the Provisions are the minimum 
requirements. It could be the decision of the owner or registered design professional to include more 
stringent quality assurance requirements. The primary method for achieving quality assurance is 
through the use of special inspectors and testing agencies. 
Registered design professional(s) in responsible charge, or their employees, may perform the special 
inspections, when approved by the authority having jurisdiction. Increased involvement by the 
registered design professional in responsible charge allows for early detection of problems during 
construction when they can be resolved more easily. 
2.2 GENERAL REQUIREMENTS 
Because of the complexity of design and construction for structures included in Seismic Design 
Categories C, D, E, and F, it is necessary to provide a comprehensive written quality assurance plan to 
assure adequate quality controls and verification during construction. Each portion of the quality 
assurance plan is required to be prepared by the registered design professional responsible for the design 
of the seismic-force-resisting system(s) and other designated seismic systems that are subject to 
requirements for quality assurance. When completed, the quality assurance plan must be submitted to 
the owner and to the authority having jurisdiction. 
The performance for quality control of the contractors and subcontractors varies from project to project. 
The quality assurance plan provides an opportunity for the registered design professional to delineate 
the types and frequency of testing and inspections, and the extent of the structural observations to be 
performed during the construction process and to assure that the construction is in conformance with the 
approved construction documents. Special attention should be given in the quality assurance plan for 
projects with higher occupancy importance factors. 
The authority having jurisdiction shall approve the quality assurance plan and shall obtain from each 
contractor a written statement that the contractor understands the requirements of the quality assurance 
plan and will exercise the necessary control to obtain conformance. The exact methods of control are 
the responsibility of the individual contractors, subject to approval by the authority having jurisdiction. 
Special inspections, in addition to those included in the quality assurance plan, may be required by the 
authority having jurisdiction to ensure that there is compliance with the approved construction 
documents. 
As indicated in Sec. 2.2, certain regular, low-rise structures assigned to Seismic Use Group I are exempt 
from preparation of a quality assurance plan. Any structure that does not satisfy all of the criteria 
included in the exception or is not otherwise exempted by the Provisions is required to have a quality 
 Quality Assurance 
15 
assurance plan. It is important to emphasize that this exemption only applies to the preparation of a 
quality assurance plan. All special inspections and testing that are otherwise required by the Provisions 
must be performed. 
2.3 SPECIAL INSPECTION 
Special inspection is the monitoring of materials and workmanship that are critical to the integrity of the 
structure. The requirements listed in this section, from foundation systems through cold-formed steel 
framing, have been included in the national model codes for many years. It is a premise of the 
Provisions that there will be an adequate supply of knowledgeable and experienced inspectors available 
to provide the necessary special inspections for the structural categories of work. Special training 
programs may have to be developed and implemented for the nonstructural categories. 
A special inspector is a person approved by the authority having jurisdiction as being qualified to 
perform special inspections for the category of work involved. As a guide to the authority having 
jurisdiction, it is suggested that the special inspector is to be one of the following: 
1. A person employed and supervised by the registered design professional in responsible charge for 
the design of the designated seismic system or the seismic-force-resisting system for which the 
special inspector is engaged. 
2. A person employed by an approved inspection and/or testing agency who is under the direct 
supervision of a registered design professional also employed by the same agency, using inspectors 
or technicians qualified by recognized industry organizations as approved by the authority having 
jurisdiction. 
3. A manufacturer or fabricator of components, equipment, or machinery that has been approved for 
manufacturing components that satisfy seismic safety standards and that maintain a quality 
assurance plan approved by authority having jurisdiction. The manufacturer or fabricator is 
required to provide evidence of such approval by means of clear marks on each designated seismic 
system or seismic-force-resisting system component shipped to the construction site. 
The extent and duration of special inspections, types of testing, and the frequency of the testing must be 
clearly delineated in the quality assurance plan. In some instances the Provisions allow periodic special 
inspection rather than continuous special inspection. Where periodic special inspections are allowed, 
the Provisions do not state specific requirements for frequency of periodic inspection, but do indicate 
stages of construction at which inspection is required for a particular category of work. The quality 
assurance plan should generally indicate the timing and extent of any periodic special inspections 
required by the Provisions. 
2.3.9 Architectural components. It is anticipated that the minimum requirements for architectural 
components (such as exterior cladding) are satisfied if the method of anchoring components and the 
number, spacing, and types of fasteners used conform to approved construction documents. 
For ceilings and access floors compliance with the construction documents should concentrate on 
critical details. For ceiling grids those details are the location and installation for grid bracing, the 
connection of runners to the perimeter edge member along two adjacent sides, and the gap provided 
between ends of runners and the edge member on the remaining two sides. 
2.3.10 Mechanical and electrical components. It is anticipated that the minimum requirements for 
mechanical and electrical components are satisfied if the method of anchoring components and the 
number, spacing, and types of fasteners actually used conform to the approved construction documents. 
It is noted that such special inspection requirements are for selected electrical, lighting, piping, and 
ductwork components in any Seismic Design Category except A or B, and for all electrical equipment in 
Seismic Design Category E or F. 
2.4 TESTING 
2 003 Commentary, Chapter 2 
16 
Compliance with nationally recognized test standards provides the authority having jurisdiction and the 
owner a means to determine the acceptability of materials and their placement. Most test standards for 
materials are developed and maintained by the American Society for Testing and Materials (ASTM). 
Through their reference in model building codes and material specifications, ASTM Standards and other 
standard testing procedures provide a uniform measure for acceptance of materials and construction. 
The Provisions and the model building codes require that standard tests be performed by an approved 
testing agency. 
Special inspector(s) are responsible for the observation and verification of the testing procedures 
performed in the field. Special inspectors determine compliance with test standards based on their 
interpretation of the standards, as measured against acceptance criteria that are included in the 
construction documents and the quality assurance plan. 
Test standards also assign responsibility to others. For example, the ASTM A 706 specification for lowalloy 
steel reinforcing bars requires the manufacturer to report the chemical composition and carbon 
equivalent of the material. In addition, the ANSI/AWS D1.4 Welding Code requires the contractor to 
prepare written specifications for the welding of reinforcing bars. It is necessary, therefore, that each 
member of the construction team has a thorough knowledge of the specified test standards that cover 
their particular work. 
2.4.5 Mechanical and electrical equipment. The registered design professional should consider 
requirements to demonstrate the seismic performance of mechanical and electrical components critical 
to the post-earthquake life safety of the occupants. Any requirements should be clearly indicated on the 
construction documents. Any currently accepted technology should be acceptable to demonstrate 
compliance with the requirements. 
It is intended that the certificate only be requested for components with an importance factor (Ip) greater 
than 1.00 and only if the component has a doubtful or uncertain seismic load path. This certificate 
should not be requested to validate functionality concerns. 
In the context of the Provisions, seismic adequacy of the component is of concern only when the 
component is required to remain operational after an earthquake or contains material that can pose a 
significant hazard if released. Meeting the requirements of this section shall be considered as an 
acceptable demonstration of the seismic adequacy of a component. 
2.5 STRUCTURAL OBSERVATIONS 
The purpose of structural observations is to allow the registered design professional(s) in responsible 
charge or other registered design professional(s) to visit the site to observe the seismic-force-resisting 
systems. Observations include verifying that the seismic-force-resisting system is constructed in general 
conformance with the construction documents, that the intent of the design has been accomplished, and 
that a complete lateral load path exists. 
Every effort shall be made to have the registered design professional in responsible charge make the 
observations. If another registered design professional performs the observations he is expected to be 
familiar with the construction documents and the design concept. 
2.6 REPORTING AND COMPLIANCE PROCEDURES 
The purpose of this section is to keep key parties informed of the special inspectors observations and 
the contractors corrections. 

 17 
Chapter 3 Commentary 
GROUND MOTION 
3.1 GENERAL 
3.1.3 Definitions. The Provisions are intended to provide uniform levels of performance for 
structures, depending on their occupancy and use and the risk to society inherent in their failure. Sec. 
1.2 of the Provisions establishes a series of Seismic Use Groups, which are used to assign each 
structure to a specific Seismic Design Category. It is the intent of the Provisions that meeting the 
seismic design criteria will provide a uniform margin against failure for all structures within a given 
Seismic Use Group. 
In past editions of the Provisions, seismic hazards around the nation were defined at a uniform 10 
percent probability of exceedance in 50 years and the design requirements were based on assigning a 
structure to a Seismic Hazard Exposure Group and a Seismic Performance Category. While this 
approach provided for a uniform likelihood throughout the nation that the design ground motion would 
not be exceeded, it did not provide for a uniform probability of failure for structures designed for that 
ground motion. The reason for this is that the rate of change of earthquake ground motion versus 
likelihood is not constant in different regions of the United States. 
The approach adopted in the Provisions is intended to provide for a uniform margin against collapse at 
the design ground motion. In order to accomplish this, ground motion hazards are defined in terms of 
maximum considered earthquake ground motions. The maximum considered earthquake ground 
motions are based on a set of rules that depend on the seismicity of an individual region. The design 
ground motions are based on a lower bound estimate of the margin against collapse inherent in 
structures designed to the Provisions. This lower bound was judged, based on experience, to 
correspond to a factor of about 1.5 in ground motion. Consequently, the design earthquake ground 
motion was selected at a ground shaking level that is 1/1.5 (2/3) of the maximum considered 
earthquake ground motion. 
For most regions of the nation, the maximum considered earthquake ground motion is defined with a 
uniform probability of exceedance of 2 percent in 50 years (return period of about 2500 years). While 
stronger shaking than this could occur, it was judged that it would be economically impractical to 
design for such very rare ground motions and that the selection of the 2 percent probability of 
exceedance in 50 years as the maximum considered earthquake ground motion would result in 
acceptable levels of seismic safety for the nation. 
In regions of high seismicity, such as coastal California, the seismic hazard is typically controlled by 
large-magnitude events occurring on a limited number of well-defined fault systems. Ground shaking 
calculated at a 2 percent probability of exceedance in 50 years would be much larger than that which 
would be expected based on the characteristic magnitudes of earthquakes on these known active faults. 
This is because these major active faults can produce characteristic earthquakes every few hundred 
years. For these regions, it is considered more appropriate to directly determine maximum considered 
earthquake ground motions based on the characteristic earthquakes of these defined faults. In order to 
provide for an appropriate level of conservatism in the design process, when this approach to 
calculation of the maximum considered earthquake ground motion is used, the median estimate of 
ground motion resulting for the characteristic event is multiplied by 1.5. 
Sec. 4.1.1 of the Provisions defines the maximum considered earthquake ground motion in terms of the 
mapped values of the spectral response acceleration at short periods, SS , and at 1 second, S1 , for Class 
B sites. These values may be obtained directly from Maps 1 through 24, respectively. A detailed 
explanation for the development of Maps 1 through 24 appears as Appendix A to this Commentary 
volume. The procedure by which these maps were created, as described above and in Appendix A, is 
2 003 Commentary, Chapter 3 
18 
also included in the Provisions under Sec 3.4 so that registered design professionals performing such 
studies may use methods consistent with those that served as the basis for developing the maps. 
3.2 GENERAL REQUIREMENTS 
3.2.2 Procedure selection. This section sets alternative procedures for determining ground shaking 
parameters for use in the design process. The design requirements generally use response spectra to 
represent ground motions in the design process. For the purposes of the Provisions, these spectra are 
permitted to be determined using either a generalized procedure in which mapped seismic response 
acceleration parameters are referred to or by site-specific procedures. The generalized procedure in 
which mapped values are used is described in Sec. 3.3. The site-specific procedure is described in Sec. 
3.4. 
3.3 GENERAL PROCEDURE 
This section provides the procedure for obtaining design site spectral response accelerations using the 
maps provided with the Provisions. Many buildings and structures will be designed using the 
equivalent lateral force procedure of Sec. 5.2, and this general procedure to determine the design 
spectral response acceleration parameters, SDS and SD1, that are directly used in that procedure. Some 
structures will be designed using the response spectrum procedure of Sec. 5.3. This section also 
provides for the development of a general response spectrum, which may be used directly in the modal 
analysis procedure, from the design spectral response acceleration parameters, SDS and SD1. 
Maps 1 and 2 respectively provide two parameters, SS and S1, based on a national seismic hazard study 
conducted by the U.S. Geological Survey. For most buildings and sites, they provide a suitably 
accurate estimate of the maximum considered earthquake ground shaking for design purposes. For 
some sites, with special soil conditions or for some buildings with special design requirements, it may 
be more appropriate to determine a site-specific estimate of the maximum considered earthquake 
ground shaking response accelerations. Sec. 3.4 provides guidance on site-specific procedures. 
SS is the mapped value, from Map 1 of the 5-percent-damped maximum considered earthquake spectral 
response acceleration, for short period structures founded on Class B, firm rock, sites. The shortperiod 
acceleration has been determined at a period of 0.2 seconds. This is because it was concluded 
that 0.2 seconds was reasonably representative of the shortest effective period of buildings and 
structures that are designed by the Provisions, considering the effects of soil compliance, foundation 
rocking, and other factors typically neglected in structural analysis. 
Similarly, S1 is the mapped value from Map 2 of the 5-percent-damped maximum considered 
earthquake spectral response acceleration at a period of 1 second on Site Class B. The spectral 
response acceleration at periods other than 1 second can typically be derived from the acceleration at 1 
second. Consequently, these two response acceleration parameters, SS and S1, are sufficient to define 
an entire response spectrum for the period range of importance for most buildings and structures, for 
maximum considered earthquake ground shaking on Class B sites. 
In order to obtain acceleration response parameters that are appropriate for sites with other 
characteristics, it is necessary to modify the SS and S1 values, as indicated in Sec.3.3.2. This 
modification is performed with the use of two coefficients, Fa and Fv, which respectively scale the SS 
and S1 values determined for firm rock sites to values appropriate for other site conditions. The 
maximum considered earthquake spectral response accelerations adjusted for Site Class effects are 
designated SMS and SM1, respectively, for short-period and 1-second-period response. As described 
above, structural design in the Provisions is performed for earthquake demands that are 2/3 of the 
maximum considered earthquake response spectra. Two additional parameters, SDS and SD1, are used to 
define the acceleration response spectrum for this design level event. These are taken, respectively, as 
2/3 of the maximum considered earthquake values, SMS and SM1, and completely define a design 
response spectrum for sites of any characteristics. 
Ground Motion 
19 
Sec. 3.5.1 provides a categorization of the various classes of site conditions, as they affect the design 
response acceleration parameters. Sec. 3.5.2 describes the steps by which sites can be classified as 
belonging to one of these Site Classes. 
3.3.2 Site coefficients and adjusted acceleration parameters. The site coefficients Fa and Fv 
presented in Provisions Tables 3.3-1 and 3.3-2 are based on the research described in the following 
paragraphs. 
It has long been recognized that the effects of local soil conditions on ground motion characteristics 
should be considered in building design. The 1989 Loma Prieta earthquake provided abundant strong 
motion data that was used extensively together with other information in introducing the site 
coefficients Fa and Fv into the 1994 Provisions. 
The amount of ground motion amplification by a soil deposit relative to bedrock depends on the wavepropagation 
characteristics of the soil, which can be estimated from measurements or inferences of 
shear-wave velocity and in turn the shear modulus for the materials as a function of the level of 
shaking. In general, softer soils with lower shear-wave velocities exhibit higher amplifications than 
stiffer soils with higher shear velocities. Increased levels of ground shaking result in increased soil 
stress-strain nonlinearity and increased soil damping which in general reduces the amplification, 
especially for shorter periods. Furthermore, for soil deposits of sufficient thickness, soil amplification 
is generally greater at longer periods than at the shorter periods. Based on the studies summarized 
below, values of the soil amplification factors (site coefficients) shown in Tables 3.3-1 and 3.3-2 were 
developed as a function of site class and level of ground shaking. Table 3.3-1 presents the short-period 
site coefficient, Fa; Table 3.3-2 presents the long-period site coefficient, Fv. As described in Sec. 3.5, 
Site Classes A through E describe progressively softer (lower shear wave velocity) soils. 
Strong-motion recordings obtained on a variety of geologic deposits during the Loma Prieta earthquake 
of October 17, 1989 provided an important empirical basis for the development of the site coefficients 
Fa and Fv. Figure C3.3.2-1 presents average response spectra of ground motions recorded on soft clay 
and rock sites in San Francisco and Oakland during the Loma Prieta earthquake. The peak acceleration 
(which plots at zero-period of the response spectra) was about 0.08 to 0.1 g at the rock sites and was 
amplified two to three times to 0.2 g or 0.3 g at the soft soil sites. The response spectral accelerations 
at short periods (~ 0.2 or 0.3 second) were also amplified on average by factors of 2 or 3. It can be 
seen in Figure C3.3.2-1 that, at longer periods between about 0.5 and 1.5 or 2 seconds, the 
amplifications of response spectra on the soft clay site relative to rock were even greater, ranging from 
about 3 to 6 times. Ground motions on stiff soil sites were also observed to be amplified relative to 
rock sites during the Loma Prieta earthquake, but by smaller factors than on soft soils. 

2 003 Commentary, Chapter 3 
20 
Figure C3.3.2-1. Average spectra recorded during the 1989 Loma Prieta earthquake in San 
Francisco Bay area at rock sites and soft soil sites (modified after Housner, 1990). 
Average amplification factors derived from the Loma Prieta earthquake data with respect to firm to 
hard rock for short-period (0.1-0.5 sec), intermediate-period (0.5-1.4 sec), mid-period (0.4-2.0 sec), 
and long-period (1.5-5.0 sec) bands, show that a short-period factor and a mid-period factor (the midperiod 
factor was later renamed the long-period factor in the NEHRP Provisions) are sufficient to 
characterize the response of the local site conditions (Borcherdt, 1994). This important result is 
consistent with the two-factor approach to response spectrum construction summarized in Figure 
C3.3.2-2. Empirical regression curves fitted to these amplification data as a function of mean shear 
wave velocity at a site are shown in Figure C3.3.2-3. 

Ground Motion 
21 
Figure C3.3.2-2. Two-factor approach to local site response. 
0
1
2
3
4
5
6
7
100 200 300 400 500 600 700 800 900 1000 1100 1200 1300 1400 
Mean Shear-Wave Velocity to 30 m (100 ft) (Vs, m/s) 
Average Spectral Amplification Factors, 
Fa and Fv 
Loma Prieta Strong-Motion Data 
Fa - 95% 
Fa = (997 / Vs )^0.36 
Fa +95% 
Fv - 95% 
Fv = (1067 / Vs )^0.64 
Fv + 95% 
Fa (0.1g) for Site Class Intervals 
Fv (0.1g) for Site Class Intervals 
Site Class E
Site Class D 
Site Class C 
Site Class B 
Firm to Hard rocks 
Gravelly soils and Soft rocks 
Stiff clays and 
Sandy soils 
Soft soils 
Figure C3.3.2-3. Short-period Fa and long-period Fv site coefficients with respect to site class B 
(firm to hard rocks) inferred as a continuous function of shear-wave velocity from empirical 
regression curves derived using Loma Prieta strong-motion recordings. The 95 percent 
confidence intervals for the ordinate to the true population regression line and the corresponding 
site coefficients in Tables 3.3-1 and 3.3-2 for 0.1g acceleration are plotted. The curves show that a 
two factor approach with a short- and a long-period site coefficient is needed to characterize the 
response of near surface deposits (modified from Borcherdt 1994). 
The curves in Figure C3.3.2-3 provide empirical estimates of the site coefficients Fa and Fv as a 
function of mean shear wave velocity for an input peak ground accelerations on rock equal to about 0.1 
g (Borcherdt, 1994; Borcherdt and Glassmoyer, 1994) The empirical amplification factors predicted by 
these curves are in good agreement with those obtained from empirical analyses of Loma Prieta data 
2 003 Commentary, Chapter 3 
22 
for soft soils by Joyner et al. (1994) shown in Figure 3.3.2-4. These short- and long-period 
amplification factors for low peak ground (rock) acceleration levels (~ 0.1 g) provided the basis for the 
values in the left-hand columns of Tables 3.3-1 and 3.3-2. Note that in Tables 3.3-1 and 3.3-2, a peak 
ground (rock) acceleration of 0.1g corresponds approximately to a response spectral acceleration on 
rock at 0.2-second period (Ss) equal to 0.25g (Table 3.3-1) and to a response spectral acceleration on 
rock at 1.0-second period (S1) equal to 0.1g (Table 3.3-2). 
Figure C3.3.2-4. Calculation of average ratios of response spectra (RRS) curves for 5 percent 
damping from records of 1989 Loma Prieta earthquake on soft soil sites. The middle curve gives 
the geometric average ratio as function of the period. The top and bottom curves show the range 
from plus to minus one standard deviation of the average of the logarithms of the ratios. The 
vertical lines show the range from plus to minus standard deviation of the logarithms of the 
ratios (Joyner et al., 1994). 
The values of Fa and Fv obtained directly from the analysis of ground motion records from the Loma 
Prieta earthquake were used to calibrate numerical one-dimensional site response analytical techniques, 
including equivalent linear as well as nonlinear programs. The equivalent linear program SHAKE 
(Schnabel et al. 1972), which had been shown in previous studies to provide reasonable predictions of 
soil amplification during earthquakes (e.g., Seed and Idriss 1982), was used extensively for this 
calibration. Seed et al. (1994) and Dobry et al. (1994) showed that the one-dimensional model 
provided a good first-order approximation to the observed site response in the Loma Prieta earthquake, 
especially at soft clay sites. Idriss (1990, 1991) used these analysis techniques to study the 
amplification of peak ground acceleration on soft soil sites relative to rock sites as a function of the 
peak acceleration on rock. Results of these studies are shown in Figure 3.3.2-5, illustrating that the 
large amplifications of peak acceleration on soft soil for low rock accelerations recorded during the 
1985 Mexico City earthquake and the 1989 Loma Prieta earthquake should tend to decrease rapidly as 
rock accelerations increases above about 0.1 g. 

Ground Motion 
23 
Figure C3.3.2-5. Relationships between maximum acceleration on rock and other local site 
conditions (Idriss, 1990, 1991). 
After calibration, these equivalent linear and nonlinear one-dimensional site response techniques were 
used to extrapolate the values of Fa and Fv to larger rock accelerations of as much as 0.4g or 0.5g. 
Parametric studies involving combinations of hundreds of soil profiles and several dozen input 
earthquake rock motions provided quantitative guidelines for extrapolation of the Loma Prieta 
earthquake results (Seed et al. 1994; Dobry et al. 1994). Figure C3.3.2-6 summarizes some results of 
these site response analyses using the equivalent linear method. This figure presents values of peak 
amplification of response spectra at long periods for soft sites (termed maximum Ratio of Response 
Spectra, RRSmax) calculated using the equivalent linear approach as a function of the plasticity index 
(PI) of the soil and the rock shear wave velocity Vr for both weak (0.1 g) and strong (0.4 g) input rock 
shaking. The effect of PI is due to the fact that soils with higher PI exhibit less stress-strain 
nonlinearity and a lower material damping (Vucetic and Dobry 1991). For peak rock acceleration = 
0.1 g, Vr = 4,000 ft/sec (1220 m/s) and PI = 50, roughly representative of San Francisco Bay area soft 
sites in the Loma Prieta earthquake, RRSmax = 4.4, which for a soil shear wave velocity of 150 m/sec 
coincides with the upper part of the range in Figure 3.3.2-3 inferred from the ground motion records. 
Note the reduction of this value of RRSmax from 4.4 to about 3.3 in Figure C3.3.2-6 when peak rock 
acceleration = 0.4 g, due to soil nonlinearity. Results such as those in Figure C3.3.2-6 provided the 
basis for the values of Fa and Fv shown in the right-most four columns of Tables 3.3-1 and 3.3-2. 

2 003 Commentary, Chapter 3 
24 
Figure C3.3.2-6. Variation of RRSmax of uniform layer of soft clay on rock from equivalent 
linear site response analyses (Dobry et al., 1994). 
Graphs and equations that provide a framework for extrapolation of Fa and Fv from Loma Prieta results 
to larger input ground motion levels continuously as a function of site conditions (shear-wave velocity) 
are shown in Figure C3.3.2-7. The site coefficients in Tables 3.3-1 and 3.3-2 are superimposed on this 
figure. These simple curves were developed to reproduce the site coefficients for site classes E and B 
and provide approximate estimates of the coefficients for the other Site Classes at various ground 
acceleration levels. The equations describing the curves indicate that the amplification at a site is 
proportional to the shear velocity ratio (impedance ratio) with an exponent that varies with the input 
ground motion level (Borcherdt, 1994). 
Ground Motion 
25 
0.1 
1.0 
10.0
100 1000 
Mean Shear-Wave Velocity to 30 m (100 ft) (v S , m/s) 
Short-Period Amplification Factor Fa 
wrt Site Class B 
I=0.1g; ma = 0.35 
I=0.3g; ma = 0.10 
I=0.2g; ma = 0.25 
I=0.4g; ma = -0.05 
Fa (0.1g) for Site Class Intervals 
Fa from Table 3.3-1 
Soft soils 
Stiff clays and 
Sandy soils Gravelly soils and 
Soft rocks 
Firm to Hard 
rocks 
Site Class E 
Site Class D 
Site Class C Site Class B 
F a = (v B / v S ) m a = ( 1050 m/s /v S ) m a 
(a) 
0.1
1 
10
100 1000 
Mean Shear-Velocity to 30 m (100 ft) (v S , m/s) 
Long-Period Amplification Factor Fv 
wrt Site Class B 
I=0.1g; mv = 0.65 
I=0.2g; mv = 0.60 
I=0.3g; mv = 0.53 
I=0.4g; mv = 0.45 
Fv (0.1g) for Site Class Intervals 
Fv from Table 3.3-2 
Soft soils 
Stiff clays and Sandy soils Gravelly soils and 
Soft rocks Firm to Hard 
rocks 
Site Class E 
Site Class D Site Class C 
Site Class B 
F v = (v B / v S ) m v = ( 1050 m/s /v S ) m v 
(b) 
Figure C3.3.2-7. Graphs and equations that provide a simple framework for inference of (a) Fa 
and (b) Fv values as a continuous function of shear velocity at various input acceleration levels. 
Site coefficients in Table 3.3-1 and 3.3-2 are superimposed. These simple curves were developed 
to reproduce the site coefficients for site classes E and B and provide approximate estimates of 
the coefficients for the other site classes at various ground acceleration levels (from Borcherdt 
1994). 
A more extensive discussion of the development of site coefficients is presented by Dobry, et al. 
(2000). Since the development of these coefficients and the development of a community consensus 
regarding their values in 1992, recent earthquakes have provided additional strong motion data from 
which to infer site amplifications. Analyses conducted on the basis of these more recent data are 
reported by a number of researchers, including Crouse and McGuire, 1996; Dobry et al., 1999; Silva et 
al., 2000; Joyner and Boore, 2000; Field, 2000; Steidl, 2000; Rodriquez-Marek et al., 2001; Borcherdt, 
2002, and Stewart et al., 2003. While the results of these studies vary, overall the site amplification 
2 003 Commentary, Chapter 3 
26 
factors are generally consistent with those in Tables 3.3-1 and 3.3.2 and there is no clear consensus for 
change at the present time (end of 2002). 
3.3.4 Design response spectrum. This section provides a general method for obtaining a 5-percentdamped 
response spectrum from the site design acceleration response parameters SaS and Sa1. This 
spectrum is based on that proposed by Newmark and Hall, as a series of three curves representing in 
the short period, a region of constant spectral response acceleration; in the long period, a range of 
constant spectral response velocity; and in the very long period, a range of constant spectral response 
displacement. Response acceleration at any period in the long period range can be related to the 
constant response velocity by the equation: 
2 
Sa Sv Sv 
T
p 
=. = (C3.3-1) 
where . is the circular frequency of motion, T is the period, and Sv is the constant spectral response 
velocity. Thus the site design spectral response acceleration at 1 second, Sa1, is simply related to the 
constant spectral velocity for the spectrum as follows: 
Sa1 = 2pSv (C3.3-2) 
and the spectral response acceleration at any period in the constant velocity range can be obtained from 
the relationship: 
D1 
a 
S S
T 
= (C3.3-3) 
The constant displacement domain of the response spectrum is not included on the generalized 
response spectrum because relatively few structures have a period long enough to fall into this range. 
Response accelerations in the constant displacement domain can be related to the constant 
displacement by a 1/T2 relationship. Sec. 5.3 of the Provisions, which provides the requirements for 
modal analysis also provides instructions for obtaining response accelerations in the very long period 
range. 
The TL maps were prepared following a two-step procedure. The first step consisted of establishing a 
correlation between earthquake magnitude and TL. This correlation was established by (1) determining 
the corner period between intermediate and long period motions based on seismic source theory and (2) 
examining the response spectra of strong motion accelerograms recorded during moderate and large 
magnitude earthquakes. This corner period, Tc, marks the transition between the constant displacement 
and constant velocity segments of the Fourier spectrum representing a theoretical fault-rupture 
displacement history. Tc, which was considered an approximation for TL, was related to moment 
magnitude, M, through the formula, log Tc = -1.25 + 0.3 M. This formula was selected from several 
available formulas based on comparisons of Tc predicted by this equation and TL estimated from strong 
motion accelerograms with reliable long period content. The results were used to establish the 
following half-unit ranges of M for given values of Tc.. 
Ground Motion 
27 
M Tc (sec) 
6.0  6.5 4 
6.5  7.0 6 
7.0  7.5 8 
7.5  8.0 12 
8.0  8.5 16 
8.5  9.0+ 20 
To determine the TL values for the U.S., the USGS constructed maps of the modal magnitudes (Md) in 
half-unit increments (as shown in the above table). The maps were prepared from a deaggregation of 
the 2 percent in 50-yr hazard for Sa (T = 2 sec), the response spectral acceleration at an oscillator period 
of 2 sec. (for HI the deaggregation was only available for T = 1 sec). The Md that was computed 
represented the magnitude interval that had the largest contribution to the 2 percent in 50-yr hazard for 
Sa. 
The Md maps were judged to be an acceptable approximation to values of Md that would be obtained if 
the deaggregation could have been computed at the longer periods of interest. These Md maps were 
color coded to more easily permit the eventual construction of the TL maps. Generally the TL maps 
corresponded to the Md maps, but some smoothing of the boundaries separating TL regions was 
necessary to make them more legible. A decision was made to limit the TL in the broad area in the 
central and eastern U.S., which had an Md of 16 sec, to 12 sec, Likewise, the TL for the areas affected 
by the great megathrust earthquakes in the Pacific Northwest and Alaska, was limited to 16 sec. 
3.4 SITE-SPECIFIC PROCEDURE 
The objective in conducting a site-specific ground motion analysis is to develop ground motions that 
are determined with higher confidence for the local seismic and site conditions than can be determined 
from national ground motion maps and the general procedure of Sec. 3.3. Accordingly, such studies 
must be comprehensive and incorporate current scientific interpretations. Because there is typically 
more than one scientifically credible alternative for models and parameter values used to characterize 
seismic sources and ground motions, it is important to formally incorporate these uncertainties in a sitespecific 
probabilistic analysis. For example, uncertainties may exist in seismic source location, extent 
and geometry; maximum earthquake magnitude; earthquake recurrence rate; choices for ground motion 
attenuation relationships; and local site conditions including soil layering and dynamic soil properties 
as well as possible two- or three-dimensional wave propagation effects. The use of peer review for a 
site-specific ground motion analysis is encouraged. 
Near-fault effects on horizontal response spectra include (1) directivity effects that increase ground 
motions for periods of vibration greater than approximately 0.5 second for fault rupture propagating 
toward the site; and (2) directionality effects that increase ground motions for periods greater than 
approximately 0.5 second in the direction normal (perpendicular) to the strike of the fault. Further 
discussion of these effects is contained in Somerville et al. (1997) and Abrahamson (2000). 
Conducting site-specific geotechnical investigations and dynamic site response analyses. 
Provisions Tables 3.3-1 and 3.3-2 and Sec. 3.5.1 require that site-specific geotechnical investigations 
and dynamic site response analysis be performed for sites having Site Class F soils. Guidelines are 
provided below for conducting site-specific investigations and site response analyses for these soils. 
These guidelines are also applicable if it is desired to conduct dynamic site response analyses for other 
site classes. 
Site-specific geotechnical investigation: For purposes of obtaining data to conduct a site response 
2 003 Commentary, Chapter 3 
28 
analysis, site-specific geotechnical investigations should include borings with sampling, standard 
penetration tests (SPTs) for sandy soils, cone penetrometer tests (CPTs), and/or other subsurface 
investigative techniques and laboratory soil testing to establish the soil types, properties, and layering 
and the depth to rock or rock-like material. For very deep soil sites, the depth of investigation need not 
necessarily extend to bedrock but to a depth that may serve as the location of input motion for a 
dynamic site response analysis (see below). It is desirable to measure shear wave velocities in all soil 
layers. Alternatively, shear wave velocities may be estimated based on shear wave velocity data 
available for similar soils in the local area or through correlations with soil types and properties. A 
number of such correlations are summarized by Kramer (1996). 
Dynamic site response analysis: Components of a dynamic site response analysis include the 
following steps: 
1. Modeling the soil profile: Typically, a one-dimensional soil column extending from the ground 
surface to bedrock is adequate to capture first-order site response characteristics. For very deep 
soils, the model of the soil columns may extend to very stiff or very dense soils at depth in the 
column. Two- or three-dimensional models should be considered for critical projects when two or 
three-dimensional wave propagation effects should be significant (e.g., in basins). The soil layers 
in a one-dimensional model are characterized by their total unit weights and shear wave velocities 
from which low-strain (maximum) shear moduli may be obtained, and by relationships defining the 
nonlinear shear stress-strain relationships of the soils. The required relationships for analysis are 
often in the form of curves that describe the variation of soil shear modulus with shear strain 
(modulus reduction curves) and by curves that describe the variation of soil damping with shear 
strain (damping curves). In a two- or three-dimensional model, compression wave velocities or 
moduli or Poisson ratios also are required. In an analysis to estimate the effects of liquefaction on 
soil site response, the nonlinear soil model also must incorporate the buildup of soil pore water 
pressures and the consequent effects on reducing soil stiffness and strength. Typically, modulus 
reduction curves and damping curves are selected on the basis of published relationships for 
similar soils (e.g., Seed and Idriss, 1970; Seed et al., 1986; Sun et al., 1988; Vucetic and Dobry, 
1991; Electric Power Research Institute, 1993; Kramer, 1996). Site-specific laboratory dynamic 
tests on soil samples to establish nonlinear soil characteristics can be considered where published 
relationships are judged to be inadequate for the types of soils present at the site. Shear and 
compression wave velocities and associated maximum moduli should be selected on the basis of 
field tests to determine these parameters or published relationships and experience for similar soils 
in the local area. The uncertainty in soil properties should be estimated, especially the uncertainty 
in the selected maximum shear moduli and modulus reduction and damping curves. 
2. Selecting input rock motions: Acceleration time histories that are representative of horizontal rock 
motions at the site are required as input to the soil model. Unless a site-specific analysis is carried 
out to develop the rock response spectrum at the site, the maximum considered earthquake (MCE) 
rock spectrum for Site Class B rock can be defined using the general procedure described in Sec. 
3.3. For hard rock (Site Class A), the spectrum may be adjusted using the site factors in 
Tables 3.3-1 and 3.3-2. For profiles having great depths of soil above Site Class A or B rock, 
consideration can be given to defining the base of the soil profile and the input rock motions at a 
depth at which soft rock or very stiff soil of Site Class C is encountered. In such cases, the MCE 
rock response spectrum may be taken as the spectrum for Site Class C defined using the site factors 
in Tables 3.3-1 and 3.3-2. Several acceleration time histories of rock motions, typically at least 
four, should be selected for site response analysis. These time histories should be selected after 
evaluating the types of earthquake sources, magnitudes, and distances that predominantly 
contribute to the seismic hazard at the site. Preferably, the time histories selected for analysis 
should have been recorded on geologic materials similar to the site class of materials at the base of 
the site soil profile during earthquakes of similar types (e.g. with respect to tectonic environment 
and type of faulting), magnitudes, and distances as those predominantly contributing to the site 
Ground Motion 
29 
seismic hazard. The U.S. Geological Survey national seismic hazard mapping project website 
(http://geohazards.cr.usgs.gov/eq/) includes hazard deaggregation options and can be used to 
evaluate the predominant types of earthquake sources, magnitudes, and distances contributing to 
the hazard. Sources of recorded acceleration time histories include the data bases of the 
Consortium of Organizations for Strong Motion Observation Systems (COSMOS) Virtual Data 
Center web site (db.cosmos-eq.org) and the Pacific Earthquake Engineering Research Center 
(PEER) Strong Motion Data Base website (http://peer.berkeley.edu/smcat/). Prior to analysis, each 
time history should be scaled so that its spectrum is at the approximate level of the MCE rock 
response spectrum in the period range of interest. It is desirable that the average of the response 
spectra of the suite of scaled input time histories be approximately at the level of the MCE rock 
response spectrum in the period range of interest. Because rock response spectra are defined at the 
ground surface rather than at depth below a soil deposit, the rock time histories should be input in 
the analysis as outcropping rock motions rather than at the soil-rock interface. 
3. Site response analysis and results interpretation: Analytical methods may be equivalent linear or 
nonlinear. Frequently used computer programs for one-dimensional analysis include the 
equivalent linear program SHAKE (Schnabel et al., 1972; Idriss and Sun, 1992) and the nonlinear 
programs DESRA-2 (Lee and Finn, 1978), MARDES (Chang et al., 1991), SUMDES (Li et al., 
1992), D-MOD (Matasovic, 1993), TESS (Pyke, 1992), and DESRAMUSC (Qiu, 1998). If the 
soil response is highly nonlinear (e.g., high acceleration levels and soft soils), nonlinear programs 
may be preferable to equivalent linear programs. For analysis of liquefaction effects on site 
response, computer programs incorporating pore water pressure development (effective stress 
analyses) must be used (e.g., DESRA-2, SUMDES, D-MOD, TESS, and DESRAMUSC). 
Response spectra of output motions at the ground surface should be calculated and the ratios of 
response spectra of ground surface motions to input outcropping rock motions should be 
calculated. Typically, an average of the response spectral ratio curves is obtained and multiplied 
by the MCE rock response spectrum to obtain a soil response spectrum. Sensitivity analyses to 
evaluate effects of soil property uncertainties should be conducted and considered in developing 
the design response spectrum. 
3.4.2 Deterministic maximum considered earthquake. It is required that ground motions for the 
deterministic maximum considered earthquake be based on characteristic earthquakes on all known 
active faults in a region. As defined in Sec. 3.1.3, the magnitude of a characteristic earthquake on a 
given fault should be a best-estimate of the maximum magnitude capable for that fault but not less than 
the largest magnitude that has occurred historically on the fault. The maximum magnitude should be 
estimated considering all seismic-geologic evidence for the fault, including fault length and 
paleoseismic observations. For faults characterized as having more than a single segment, the potential 
for rupture of multiple segments in a single earthquake should be considered in assessing the 
characteristic maximum magnitude for the fault. 
3.5 SITE CLASSIFICATION FOR SEISMIC DESIGN 
3.5.1 Site Class Definitions. Based on the studies and observations discussed in Sec. 3.3-2, the site 
categories in the 2003 Provisions are defined in terms of the small-strain shear wave velocity in the top 
100 ft (30 m) of the profile, v s , as might be inferred from travel time for a shear wave to travel from 
the surface to a depth of 100 ft (30m). If shear wave velocities are available for the site, they should be 
used to classify the site. 
However, in recognition of the fact that in many cases the shear wave velocities are not available, 
alternative definitions of the site classes also are included in the 2003 Provisions. They use the 
standard penetration resistance for cohesionless and cohesive soils and rock and the undrained shear 
strength for cohesive soils only. These alternative definitions are rather conservative since the 
correlation between site amplification and these geotechnical parameters is more uncertain than the 
correlation with v s . That is, there will be cases where the values of Fa and Fv will be smaller if the site 
2 003 Commentary, Chapter 3 
30 
category is based on v s rather than on the geotechnical parameters. Also, the site category definitions 
should not be interpreted as implying any specific numerical correlation between shear-wave velocity 
and standard penetration resistance or shear strength. 
Equation 3.5-1 is for inferring the average shear-wave velocity to a depth of 100 ft (30m) at a site. 
Equation 3.5-1specifies that the average velocity is given by the sum of the thicknesses of the geologic 
layers in the upper 100 ft divided by the sum of the times for a shear wave to travel through each layer, 
where travel time for each layer is specified by the ratio of the thickness and the shear wave velocity 
for the layer. It is important that this method of averaging be used as it may result in a significantly 
lower effective average shear wave velocity than the velocity that would be obtained by averaging the 
velocities of the individual layers directly. 
Equation 3.5-2 is for classifying the site using the standard penetration resistance (N-value) for 
cohesionless soils, cohesive soils, and rock in the upper 100 ft (30 m). A method of averaging 
analogous to the method of Equation 3.5-1 for shear wave velocity is used. The maximum value of N 
that can be used for any depth of measurement in soil or rock is 100 blows/ft. 
Equations 3.5-3 and 3.5-4 are for classifying the site using the standard penetration resistance of 
cohesionless soil layers, Nch, and the undrained shear strength of cohesive soil layers, su, within the top 
100 ft (30 m). These equations are provided as an alternative to using Eq. 3.5-2 for which N-values in 
all geologic materials in the top 100 ft (30 m) are used. When using Eq. 3.5-3 and 3.5-4, only the 
thicknesses of cohesionless soils and cohesive soils within the top 100 ft (30 m) are used. 
As indicated in Sec. 3.3-2 and 3.5-1, soils classified as Site Class F according to the definitions in Sec. 
3.5-1 require site-specific evaluations. An exception is made, however, for liquefiable sites where the 
structure has a fundamental period of vibration equal to or less than 0.5 second. For such structures, 
values of Fa and Fv for the site may be determined using the site class definitions and criteria in Sec. 
3.5-1 assuming liquefaction does not occur. The exception is provided because ground motion data 
obtained in liquefied soil areas during earthquakes indicate that short-period ground motions are 
generally attenuated due to liquefaction whereas long-period ground motions may be amplified. This 
exception is only for the purposes of defining the site class and obtaining site coefficients. It is still 
required to assess liquefaction potential and its effects on structures as a ground failure hazard as 
specified in Chapter 7. 
3.5.2 Steps for classifying a site. A step- by- step procedure for classifying a site is given in the 
Provisions. Although the procedure and criteria in Sec. 3.5.1 and 3.5.2 are straightforward, there are 
aspects of these assessments that may require additional judgment and interpretation. Highly variable 
subsurface conditions beneath a building footprint could result in overly conservative or 
unconservative site classification. Isolated soft soil layers within an otherwise firm soil site may not 
affect the overall site response if the predominant soil conditions do not include such strata. 
Conversely, site response studies have shown that continuous, thin, soft clay strata may increase the 
site amplification. 
The site class should reflect the soil conditions that will affect the ground motion input to the structure 
or a significant portion of the structure. For structures receiving substantial ground motion input from 
shallow soils (e.g. structures with shallow spread footings, laterally flexible piles, or structures with 
basements where it is judged that substantial ground motion input to the structure may come through 
the side walls), it is reasonable to classify the site on the basis of the top 100 ft (30 m) of soils below 
the ground surface. Conversely, for structures with basements supported on firm soils or rock below 
soft soils, it is reasonable to classify the site on the basis of the soils or rock below the mat, if it can be 
justified that the soft soils contribute very little to the response of the structure. 
Buildings on sloping bedrock sites and/or having highly variable soil deposits across the building area 
require careful study since the input motion may vary across the building (for example, if a portion of 
the building is on rock and the rest is over weak soils). Site-specific studies including two- or three-
Ground Motion 
31 
dimensional modeling may be appropriate in such cases to evaluate the subsurface conditions and site 
and superstructure response. Other conditions that may warrant site-specific evaluation include the 
presence of low shear wave velocity soils below a depth of 100 ft (30 m), location of the site near the 
edge of a filled-in basin, or other subsurface or topographic conditions with strong two- and threedimensional 
site-response effects. Individuals with appropriate expertise in seismic ground motions 
should participate in evaluations of the need for and nature of such site-specific studies. 
REFERENCES 
Abrahamson, N.A., 2000, Effects of rupture directivity on probabilistic seismic hazard analysis: 
Proceedings, Sixth International Conference on Seismic Zonation, Palm Springs, California. 
Borcherdt, R.D., 2002, Empirical evidence for site coefficients in building-code provisions, Earthquake 
Spectra, Earthquake Engineering Research Institute, v. 18, No. 2, p. 189-217. 
Borcherdt, R.D., 1994, Estimates of site-dependent response spectra for design (methodology and 
justification), Earthquake Spectra, 10, 617-653, (Simplified site classes and empirical amplification 
factors for site dependent code provisions, Proceedings of the 1992 NCEER/SEAOC/BSSC Workshop 
on Site Response During Earthquakes and Seismic Code Provisions, G.R. Martin, ed., University of 
Southern California, Los Angeles, November 18-20, 1992, National Center for Earthquake Engineering 
Research Special Publication NCEER-94-SP01, Buffalo, New York). 
Borcherdt, R.D., and Glassmoyer, G., 1994, Influences of local geology on strong and weak ground 
motions in the San Francisco Bay region, California, and their implications for site-specific code 
provisions, in The Loma Prieta Earthquake of October 17, 1989  Strong Ground Motion, R.D. 
Borcherdt, ed., U.S. Geological Survey Professional Paper 1551-A, A77-A108. 
Chang, C.-Y., C. M. Mok, M. S. Power, and Y. K. Tang. 1991. Analysis of Ground Response at 
Lotung Large-Scale Soil-Structure Interaction Experiment Site, Report NP-7306-SL. Palo Alto, 
California: Electric Power Research Institute. 
Crouse, C.B., and McGuire, J.W., 1996, Site response studies for purposes of revising NEHRP Seismic 
Provisions, Earthquake Spectra, Earthquake Engineering Research Institute, v. 12, no. 3. 
Dobry, R., Ramos, R., and Power, M.S., 1999, Site factors and site categories in seismic codes, 
Technical Report MCEER-99-0010, Multidisciplinary Center for Earthquake Engineering Research. 
Dobry, R., Borcherdt, R., Crouse, C.B., Idriss, I.M., Joyner, W.B., Martin, G.R., Power, M.S., Rinne, 
E.E., Seed, R.B., 2000, New Site coefficients and site classifications system used in recent building 
seismic code provisions, Earthquake Spectra 16 (1); 41-67. 
Dobry, R., Martin, G. R., Parra, E., and Bhattacharyya, A., 1994, Development of site-dependent ratios 
of elastic response spectra (RRS) and site categories for building seismic codes, Proceedings of the 
1992 NCEER/SEAOC/BSSC Workshop on Site Response During Earthquakes and Seismic Code 
Provisions, G.R. Martin, ed., University of Southern California, Los Angeles, November 18-20, 1992, 
National Center for Earthquake Engineering Research Special Publication NCEER-94-SP01, Buffalo, 
NY. 
Electric Power Research Institute. 1993. Guidelines for Determining Design Basis Ground Motions, 
Report EPRI TR-102293. Palo Alto, California. 
Field, E.H., 2000, A modififed ground motion attenuation relationship for southern California that 
accounts for detailed site classification and a basin depth effect, Bulletin of the Seismological Society of 
America, 90, S209-S221. 
Housner, G.W., 1990, Competing against time: Report to Governor George Duekmejian from the 
Governers Board of Inquiry on the 1989 Loma Pireta earthquake, May. 
Idriss, I. M. 1990, Response of Soft Soil Sites During Earthquakes. In Proceedings of the 
2 003 Commentary, Chapter 3 
32 
Symposium to Honor Professor H. B. Seed, Berkeley, May, 273-289. 
Idriss, I.M., 1991, Earthquake ground motions at soft soil sites, Proceedings of the Second 
InternationalConference on Recent Advances in Geotechnical Earthquake Engineering and Soil 
Dynamics, St. Louis, MO, III, 2265-2273. 
Idriss, I. M., and Sun, J.I., 1992. User's Manual for SHAKE91. Davis: Center for Geotechnical 
Modeling, Department of Civil and Environmental Engineering, University of California. 
Joyner, W.B., and Boore, D.M., 2000, Recent developments in earthquake ground motion estimation, 
Proceeding 6th International Conference on Seismic Zonation, Palm Springs. 
Joyner, W.B., Fumal, T.E., and Glassmoyer, G., 1994, Empirical spectral response ratios for strong 
motion data from the 1989 Loma Prieta, California earthquake, Proceedings for the 1992 
NCEER/SEAOC/BSSC Workshop on Site Response During Earthquakes and Seismic Code Provisions, 
G.R. Martin, ed., University of Southern California, Los Angeles, November 18-20, 1992, National 
Center for Earthquake Engineering Research Special Publication NCEER-94-SP01, Buffalo, NY. 
Kramer, S. L., 1996. Geotechnical Earthquake Engineering. New Jersey: Prentice Hall. 
Lee, M. K. W., and W. D. L. Finn. 1978. DESRA-2, Dynamic Effective Stress Response Analysis of 
Soil Deposits with Energy Transmitting Boundary Including Assessment of Liquefaction Potential, Soil 
Mechanics Series 36. Vancouver, Canada: Department of Civil Engineering, University of British 
Columbia. 
Li, X. S., Z. L. Wang, and C. K. Shen. 1992. SUMDES, A Nonlinear Procedure for Response 
Analysis of Horizontally-Layered Sites Subjected to Multi-Directional Earthquake Loading. Davis: 
Department of Civil Engineering, University of California. 
Matasovic, N. 1993. Seismic Response of Composite Horizontally-Layered Soil Deposits, Ph.D. 
dissertation, Civil and Environmental Engineering Department, University of California, Los Angeles. 
Qiu, P., 1998, Earthquake-induced nonlinear ground deformation analyses: Ph.D. dissertation, 
University of Southern California, Los Angeles. 
Pyke, R. M. 1992. TESS: A Computer Program for Nonlinear Ground Response Analyses. 
Lafayette, California: TAGA Engineering Systems and Software. 
Rodriguez-Marek, A., Bray, J.D., Abrahamson, N., 2001, An empirical geotechnical site response 
procedure, Earthquake Spectra, Earthquake Engineering Research Institute, v. 17, No. 1, p. 65-87. 
Schnabel, P.B., Lysmer, J., and Seed, H.B., 1972, SHAKE: A computer program for earthquake 
response analysis of horizontally layered sites, Report EERC 72-12, Earthquake Engineering Research 
Center, University of California, Berkeley, CA. 
Seed, R. B. Dickenson, S.E., Rau, G.A., White, R.K., and Mok, C.M., 1994, Observations regarding 
seismic response analyses for soft and deep clay sites, Proceedings of the 1992 NCEER/SEAOC/BSSC 
Workshop on Site Response During Earthquakes and Seismic Code Provisions, G.R. Martin, ed., 
University of Southern California, Los Angeles, November 18-20, 1992, National Center for 
Earthquake Engineering Research Special Publication NCEER-94-SP01, Buffalo, NY. 
Seed, H.B., and Idriss, I.M., 1982, Ground motions and soil liquefaction during earthquakes, EERI 
Monograph Series, Earthquake Engineering Research Institute, Berkeley, CA. 
Seed, H.B., R. T. Wong, I. M. Idriss, and K. Tokimatsu. 1986. Moduli and Damping Factors for 
Dynamic Analyses of Cohesionless Soils, Journal of Geotechnical Engineering, ASCE, Vol. 112, No. 
11: 1016-1032. 
Seed, H.B., and I. M. Idriss. 1970, Soil Moduli and Damping Factors for Dynamic Response Analyses, 
Report EERC 70-10. Berkeley: University of California, Earthquake Engineering Research Center. 
Ground Motion 
33 
Silva, W., Darragh, R., Gregor, N., Martin, G., Abrahamson, N., and Kircher, C., 2000, Reassessment 
of site coefficients and near-fault factors for building code provisions; Program Element: II, 98-HQGR-
1010, Report to USGS. 
Somerville, P.G., Smith, N.F., Graves, R.W., and Abrahamson, N.A., 1997, Modification of empirical 
strong ground motion attenuation relations to include the amplitude and duration effects of rupture 
directivity: Seismological Research Letters, v. 68, p. 199-222. 
Stewart, J.P, Liu, A.H., and Choi, Y., 2003, Amplification factors for spectral acceleration in 
tectonically active regions, Bulletin of the Seismological Society of America, 93(1), 332-352. 
Steidl, J.H., 2000, Site response is southern California for probabilistic seismic hazard analysis, 
Bulletin of the Seismological Society of America, 90, S149-S169. 
Sun, J. I., R. Golesorkhi, and H. B. Seed. 1988. Dynamic Moduli and Damping Ratios for Cohesive 
Soils, Report UBC/EERC-88/15. Berkeley: University of California, Earthquake Engineering 
Research Center. 
Vucetic, M., and R. Dobry. 1991. Effect of Soil Plasticity on Cyclic Response, Journal of 
Geotechnical Engineering, ASCE, Vol. 117, No. 1: 89-107. 
2 003 Commentary, Chapter 3 
34 
This page is intentionally left blank. 
 35 
Chapter 4 Commentary 
STRUCTURAL DESIGN CRITERA 
4.1 GENERAL 
4.1.2 References. ASCE 7 is referenced for the combination of earthquake loadings with other loads as 
well as for the computation of other loads; it is not referenced for the computation of earthquake loads. 
4.2 GENERAL REQUIREMENTS 
4.2.1 Design basis. Structural design for acceptable seismic resistance includes: 
1. The selection of gravity- and seismic-force-resisting systems that are appropriate to the anticipated 
intensity of ground shaking; 
2. Layout of these systems such that they provide a continuous, regular, and redundant load path capable 
of ensuring that the structures act as integral units in responding to ground shaking; and 
3. Proportioning the various members and connections such that adequate lateral and vertical strength 
and stiffness is present to limit damage in a design earthquake to acceptable levels. 
In the Provisions, the proportioning of structural elements (sizing of individual members, connections, 
and supports) is typically based on the distribution of internal forces computed based on linear elastic 
response spectrum analyses using response spectra that are representative of, but substantially reduced 
from the anticipated design ground motions. As a result, under the severe levels of ground shaking 
anticipated for many regions of the nation, the internal forces and deformations produced in most 
structures will substantially exceed the point at which elements of the structures start to yield or buckle 
and behave in an inelastic manner. This approach can be taken because historical precedent and the 
observation of the behavior of structures that have been subjected to earthquakes in the past demonstrates 
that if suitable structural systems are selected and structures are detailed with appropriate levels of 
ductility, regularity, and continuity, it is possible to perform an elastic design of structures for reduced 
forces and still achieve acceptable performance. Therefore, these procedures adopt the approach of 
proportioning structures such that under prescribed design lateral forces that are significantly reduced, by 
the response modification coefficient R, from those that would actually be produced by a design 
earthquake they will not deform beyond a point of significant yield. The elastic deformations calculated 
under these reduced design forces are then amplified, by the deflection amplification factor Cd to estimate 
the expected deformations likely to be experienced in response to the design ground motion. (Use of the 
deflection amplification factor is specified in Sec. 5.2.6.1.) Considering the intended structural 
performance and acceptable deformation levels, Sec. 4.5.1 prescribes the story drift limits for the 
expected (amplified) deformations. These procedures differ from those in earlier codes and design 
provisions wherein the drift limits were treated as a serviceability check. 
The term significant yield is not the point where first yield occurs in any member but, rather, is defined 
as that level causing complete plastification of at least the most critical region of the structure (such as 
formation of a first plastic hinge in the structure). A structural steel frame comprising compact members 
is assumed to reach this point when a plastic hinge develops in the most highly stressed member of the 
structure. A concrete frame reaches significant yield when at least one of the sections of its most highly 
stressed component reaches its strength as set forth in Chapter 9. These requirements contemplate that the 
design includes a seismic-force-resisting system with redundant characteristics wherein significant 
structural overstrength above the level of significant yield can be obtained by plastification at other points 
in the structure prior to the formation of a complete mechanism. For example, Figure C4.2-1 shows the 
2 003 Commentary, Chapter 4 
36 
lateral load-deflection curve for a typical structure. Significant yield is the level where plastification 
occurs 
at the most heavily loaded element in the structure, shown as the lowest yield hinge on the load-deflection 
diagram. With increased loading, causing the formation of additional plastic hinges, the capacity 
increases (following the solid curve ) until a maximum is reached. The overstrength capacity obtained by 
this continued inelastic action provides the reserve strength necessary for the structure to resist the 
extreme motions of the actual seismic forces that may be generated by the design ground motion. 
It should be noted that the structural 
overstrength described above results from 
the development of sequential plastic 
hinging in a properly designed, redundant 
structure. Several other sources will 
further increase structural overstrength. 
First, material overstrength (that is, actual 
material strengths higher than the nominal 
material strengths specified in the design) 
may increase the structural overstrength 
significantly. For example, a recent 
survey shows that the mean yield strength 
of A36 steel is about 30 to 40 percent 
higher than the minimum specified 
strength, which is used in design 
calculations. Second, member design 
strengths usually incorporate a strength 
reduction (or resistance) factor, f, to 
ensure a low probability of failure under 
design loading. Third, designers themselves introduce additional overstrength by selecting sections or 
specifying reinforcing patterns that exceed those required by the computations. Similar situations occur 
when minimum requirements of the Provisions, for example, minimum reinforcement ratios, control the 
design. Finally, the design of many flexible structural systems, such as moment resisting frames, are 
often controlled by the drift rather than strength limitations of the Provisions, with sections selected to 
control lateral deformations rather than provide the specified strength. The results is that structures 
typically have a much higher lateral resistance than specified as a minimum by the Provisions and first 
actual significant yielding of structures may occur at lateral load levels that are 30 to 100 percent higher 
than the prescribed design seismic forces. If provided with adequate ductile detailing, redundancy, and 
regularity, full yielding of structures may occur at load levels that are two to four times the prescribed 
design force levels. 
Figure C4.2-1 indicates the significance of design parameters contained in the Provisions including the 
response modification coefficient, R, the deflection amplification factor, Cd, and the structural 
overstrength coefficient O0. The values of the response modification coefficient, R, structural 
overstrength coefficient, O0, and the deflection amplification factor, Cd, provided in Table 4.3-1, as well 
as the criteria for story drift, including P-delta effects, have been established considering the 
characteristics of typical properly designed structures. If excessive optimization of a structural design 
is performed, with lateral resistance provided by only a few elements, the successive yield hinge behavior 
depicted in Figure C4.2-1 will not be able to form and the values of the design parameters contained in 
the Provisions may not be adequate to provide the intended seismic performance. 
The response modification coefficient, R, essentially represents the ratio of the forces that would develop 
under the specified ground motion if the structure had an entirely linearly elastic response to the 
prescribed design forces (see Figure C4.2-1). The structure is to be designed so that the level of 
significant yield exceeds the prescribed design force. The ratio R, expressed by the equation: 
Lateral Deformation (Drift), D 
succesive 
yield hinges 
Elastic response of structure 
VE 
DE 
VS Design force level 
Fully yielded strength 
DS 
Design drift 
Cd 
VY 
D 
V 
Figure C4.2-1 Inelastic force-deformation curve. 
 Structural Design Criteria 
37 
E
S 
R V
V 
= (C4.2-1) 
is always larger then 1.0; thus, all structures are designed for forces smaller than those the design ground 
motion would produce in a structure with completely linear-elastic response. This reduction is possible 
for a number of reasons. As the structure begins to yield and deform inelastically, the effective period of 
response of the structure tends to lengthen, which for many structures, results in a reduction in strength 
demand. Furthermore, the inelastic action results in a significant amount of energy dissipation, also 
known as hysteretic damping, in addition to the viscous damping. The combined effect, which is also 
known as the ductility reduction, explains why a properly designed structure with a fully yielded strength 
(Vy in Figure C4.2-1) that is significantly lower than the elastic seismic force demand (VE in Figure 
C4.2-1) can be capable of providing satisfactory performance under the design ground motion excitations. 
Defining a system ductility reduction factor Rd as the ratio between VE and VY (Newmark and Hall, 1981): 
E 
d 
Y 
R V
V 
= (C4.2-2) 
then it is clear from Figure C4.2-1 that the response modification coefficient, R, is the product of the 
ductility reduction factor and structural overstrength factor (Uang, 1991): 
R=RdO0 (C4.2-3) 
The energy dissipation resulting from hysteretic behavior can be measured as the area enclosed by the 
force-deformation curve of the structure as it experiences several cycles of excitation. Some structures 
have far more energy dissipation capacity than do others. The extent of energy dissipation capacity 
available is largely dependent on the amount of stiffness and strength degradation the structure undergoes 
as it experiences repeated cycles of inelastic deformation. Figure C4.2-2 indicates representative loaddeformation 
curves for two simple substructures, such as a beam-column assembly in a frame. Hysteretic 
curve (a) in the figure is representative of the behavior of substructures that have been detailed for ductile 
behavior. The substructure can maintain nearly all of its strength and stiffness over a number of large 
cycles of inelastic deformation. The resulting force-deformation loops are quite wide and open, 
resulting in a large amount of energy dissipation capacity. Hysteretic curve (b) represents the behavior of 
a substructure that has not been detailed for ductile behavior. It rapidly loses stiffness under inelastic 
deformation and the resulting hysteretic loops are quite pinched. The energy dissipation capacity of such 
a substructure is much lower than that for the substructure (a). Structural systems with large energy 
dissipation capacity have larger Rd values, and hence are assigned higher R values, resulting in design for 
lower forces, than systems with relatively limited energy dissipation capacity. 
Figure C4.2-2 Typical hysteretic curves. 
Some contemporary building codes, including those adopted in Canada and Europe have attempted to 
directly quantify the relative contribution of overstrength and inelastic behavior to the permissible 
2 003 Commentary, Chapter 4 
38 
reduction in design strength. Recently, the Structural Engineers Association of California proposed such 
an approach for incorporation into the 1997 Uniform Building Code. That proposal incorporated two R 
factor components, termed R0 and Rd, to represent the reduction due to structural overstrength and 
inelastic behavior, respectively. The design forces are then determined by forming a composite R, equal 
to the product of the two components (see Eq. C4.2-3). A similar approach was considered for adoption 
into the 1997 NEHRP Provisions. However, this approach was not taken for several reasons. While it 
was acknowledged that both structural overstrength and inelastic behavior are important contributors to 
the R coefficients and that they can be quantified for individual structures, it was felt that there was 
insufficient research available at the current time to support implementation in the Provisions. In 
addition, there was concern that there can be significant variation between structures in the relative 
contribution of overstrength and inelastic behavior and that, therefore, this would prevent accurate 
quantification on a system-by-system basis. Finally, it was felt that this would introduce additional 
complexity into the Provisions. While it was decided not to introduce the split R value concept into the 
Provisions in the 1997 update cycle, this should be considered in the future as additional research on the 
inelastic behavior of structures becomes available and as the sophistication of design offices improves to 
the point that quantification of structural overstrength can be done as a routine part of the design process. 
As a first step in this direction, however, the factor O0 was added to Table 4.3-1, to replace the previous 
2R/5 factor used for evaluation of brittle structural behavior modes in previous editions of the Provisions. 
The R values, contained in the current Provisions, are largely based on engineering judgment of the 
performance of the various materials and systems in past earthquakes. The values of R must be chosen 
and used with careful judgment. For example, lower values must be used for structures possessing a low 
degree of redundancy wherein all the plastic hinges required for the formation of a mechanism may be 
formed essentially simultaneously and at a force level close to the specified design strength. This 
situation can result in considerably more detrimental P-delta effects. Since it is difficult for individual 
designers to judge the extent to which R factors should be adjusted based on the inherent redundancy of 
their designs, a coefficient, ., which is calculated based on the amount of the total lateral force resisted by 
any individual element, is found in Provisions in Sec. 4.3.3. Additional discussion of this issue is 
contained in that section. 
In a departure from previous editions of the Provisions, the 1997 edition introduced an importance factor 
I into the base shear equation, which factor varies for different types of occupancies. This importance 
factor has the effect of adjusting the permissible response modification factor, R, based on the desired 
seismic performance for the structure. It recognizes that greater levels of inelastic behavior, correspond to 
increased structural damage. Thus, introducing the importance factor, I, allows for a reduction of the R 
value to an effective value R/I as a partial control on the amount of damage experienced by the structure 
under a design earthquake. Strength alone is not sufficient to obtain enhanced seismic performance. 
Therefore, the improved performance characteristics desired for more critical occupancies are also 
obtained through application of the design and detailing requirements set forth in Sec. 4.6 for each 
Seismic Design Category and the more stringent drift limits in Table 4.5-1. These factors, in addition to 
strength, are extremely important to obtaining the seismic performance desired for buildings in some 
Seismic Use Groups. 
Sec. 4.2.1 in effect calls for the seismic design to be complete and in accordance with the principles of 
structural mechanics. The loads must be transferred rationally from their point of origin to the final 
points of resistance. This should be obvious but it often is overlooked by those inexperienced in 
earthquake engineering. 
Design consideration should be given to potentially adverse effects where there is a lack of redundancy. 
Because of the many unknowns and uncertainties in the magnitude and characteristics of earthquake 
loading, in the materials and systems of construction for resisting earthquake loadings, and in the methods 
of analysis, good earthquake engineering practice has been to provide as much redundancy as possible in 
the seismic-force-resisting system of buildings. 
 Structural Design Criteria 
39 
Redundancy plays an important role in determining the ability of the building to resist earthquake forces. 
In a structural system without redundant components, every component must remain operative to preserve 
the integrity of the building structure. On the other hand, in a highly redundant system, one or more 
redundant components may fail and still leave a structural system that retains its integrity and can 
continue to resist lateral forces, albeit with diminished effectiveness. 
Redundancy often is accomplished by making all joints of the vertical load-carrying frame moment 
resisting and incorporating them into the seismic-force-resisting system. These multiple points of 
resistance can prevent a catastrophic collapse due to distress or failure of a member or joint. (The 
overstrength characteristics of this type of frame were discussed earlier in this section.) 
The designer should be particularly aware of the proper selection of R when using only one or two onebay 
rigid frames in one direction for resisting seismic loads. A single one-bay frame or a pair of such 
frames provides little redundancy so the designer may wish to consider a modified (smaller) R to account 
for a lack of redundancy. As more one-bay frames are added to the system, however, overall system 
redundancy increases. The increase in redundancy is a function of frame placement and total number of 
frames. 
Redundant characteristics also can be obtained by providing multiple different types of seismic-forceresisting 
systems in a building. The backup system can prevent catastrophic effects if distress occurs in 
the primary system. 
In summary, it is good practice to incorporate redundancy into the seismic-force-resisting system and not 
to rely on any system wherein distress in any member may cause progressive or catastrophic collapse. 
4.2.2 Combination of load effects. The load combination statements in the Provisions combine the 
effects of structural response to horizontal and vertical ground accelerations. They do not show how to 
combine the effect of earthquake loading with the effects of other loads. For those combinations, the user 
is referred to ASCE 7. The pertinent combinations are: 
1.2D + 1.0E + 0.5L + 0.2S (Additive) 
0.9D + 1.0E (Counteracting) 
where D, E, L, and S are, respectively, the effects of dead, earthquake, live, and snow loads. 
The design basis expressed in Sec. 4.2.1 reflects the fact that the specified earthquake loads are at the 
design level without amplification by load factors; thus, for sufficiently redundant structures, a load factor 
of 1.0 is assigned to the earthquake load effects in Eq. 4.2-1 and 4.2-2. 
4.2.2.1 Seismic load effect. In Eq. 4.2-1 and 4.2-2 , a factor of 0.2SDS was placed on the dead load to 
account for the effects of vertical acceleration. The 0.2SDS factor on dead load is not intended to represent 
the total vertical response. The concurrent maximum response of vertical accelerations and horizontal 
accelerations, direct and orthogonal, is unlikely and, therefore, the direct addition of responses was not 
considered appropriate. 
The . factor was introduced into Eq. 4.2-1 and 4.2-2 in the 1997 Provisions. This factor, determined in 
accordance with Sec. 4.3.3, relates to the redundancy inherent in the seismic-force-resisting system and is, 
in essence, a reliability factor, penalizing designs which are likely to be unreliable due to concentration of 
the structure=s resistance to lateral forces in a relatively few elements. 
There is very little research that speaks directly to the merits of redundancy in buildings for seismic 
resistance. The SAC joint venture recently studied the relationships between damage to welded steel 
moment frame connections and redundancy (Bonowitz et al., 1995). While this study found no specific 
correlation between damage and the number of bays of moment resisting framing per moment frame, it 
did find increased rates of damage in connections that resisted loads for larger floor areas. This study 
included modern low-, mid-, and high-rise steel buildings. 
Another study (Wood, 1991) that addresses the potential effects of redundancy evaluated the performance 
of 165 Chilean concrete buildings ranging in height from 6 to 23 stories. These concrete shear wall 
2 003 Commentary, Chapter 4 
40 
buildings with non-ductile details and no boundary elements experienced moderately strong shaking 
(MMI VII to VIII) with a strong shaking duration of over 60 seconds, yet performed well. One plausible 
explanation for this generally good performance was the substantial amount of wall area (2 to 4 percent of 
the floor area) commonly used in Chile. However, Wood=s study found no correlation between damage 
rates and higher redundancy in buildings with wall areas greater than 2 percent. 
4.2.2.2 Seismic load effect with overstrength. The seismic load effect with overstrength of Sec. 4.2.2.2 
is intended to address those situations where failure of an isolated, individual, brittle element can result in 
the loss of a complete seismic-force-resisting system or in instability and collapse. This section has 
evolved over several editions. In the 1991 Edition, a factor equal to 2R/5 factor was introduced to better 
represent the behavior of elements sensitive to overstrength in the remainder of the seismic-force-resisting 
system or in other specific structural components. The particular number was selected to correlate with 
the 3Rw/8 factor that had been introduced in the Structural Engineers Association of California (SEAOC) 
recommendations and the Uniform Building Code. This is a somewhat arbitrary factor that attempts to 
quantify the maximum force that can be delivered to sensitive elements based on historic observation that 
the real force that could develop in a structure may be 3 to 4 times the design levels. In the 1997 
Provisions, an attempt was made to determine this force more rationally through the assignment of the O0 
factor in Table 4.3-1, dependent on the individual system. Through the use of the Oo coefficient, this 
special equation provides an estimate of the maximum forces likely to be experienced by an element. 
In recent years, a number of researchers have investigated the factors that permit structures designed for 
reduced forces to survive design earthquakes. Although these studies have principally been focused on 
the development of more reliable response modification coefficients, R, they have identified the 
importance of structural overstrength and identified a number of sources of such overstrength. This has 
made it possible to replace the single 2R/5 factor formerly contained in the Provisions with a more 
system-specific estimate, represented by the O0 coefficient. 
It is recognized, that no single value, whether obtained by formula related to the R factor or otherwise 
obtained will provide a completely accurate estimate for the overstrength of all structures with a given 
seismic-force-resisting system. However, most structures designed with a given seismic-force-resisting 
system will fall within a range of overstrength values. Since the purpose of the O0 factor in Eq. 4.2-3 and 
4.2-4 is to estimate the maximum force that can be delivered to a component that is sensitive to 
overstress, the values of this factor tabulated in Table 4.3-1 are intended to be representative of the larger 
values in this range for each system. 
Figure C4.2-3 and the following discussion explore some of the factors that contribute to structural 
overstrength. The figure shows a plot of lateral structural strength vs. displacement for an elasticperfectly-
plastic structure. In addition, it shows a similar plot for a more representative real structure, that 
posses significantly more strength than the design strength. This real strength is represented by the lateral 
force Fn. Essentially, the O0 coefficient is intended to be a somewhat conservative estimate of the ratio of 
Fn to the design strength FE/R. As shown in the figure, there are three basic components to the 
overstrength. These are the design overstrength (OD), the material overstrength (OM) and the system 
overstrength (OS). Each of these is discussed separately. The design overstrength (OD) is the most 
difficult of the three to estimate. It is the difference between the lateral base shear force at which the first 
significant yield of the structure will occur (point 1 in the figure) and the minimum specified force given 
by FE/R. To some extent, this is system dependent. Systems that are strength controlled, such as most 
braced frames and shear wall structures, will typically have a relatively low value of design overstrength, 
as most designers will seek to optimize their designs and provide a strength that is close to the minimum 
specified by the Provisions. For such structures, this portion of the overstrength coefficient could be as 
low as 1.0. 
 Structural Design Criteria 
41 
Lateral Force 
Lateral Displacement (Drift) 
Elastic Response 
Elastic response force FE 
Elastic response drift 
dE 
FE/R Design force 
dE/R CddE/R 
Design drift 
Inadequate 
Strength 
2 
3 
i 
n 
F2 
Fn 
Design Overstrength 
Material Overstrength 
System Overstrength 
1 
Figure C4.2-3 Factors affecting overstrength. 
Drift controlled systems such as moment frames, however, will have substantially larger design 
overstrengths since it will be necessary to oversize the sections of such structures in order to keep the 
lateral drifts within prescribed limits. In a recent study of a number of special moment resisting steel 
frames conducted by the SAC Joint Venture design overstrengths on the order of a factor of two to three 
were found to exist (Analytical Investigation of Buildings Affected by the 1994 Northridge Earthquake, 
Volumes 1 and 2, SAC 95-04A and B. SAC Joint Venture, Sacramento, CA, 1995). Design overstrength 
also has the potential to be regionally dependent. The SAC study was conducted for frames in Seismic 
Design Categories D and E, which represent the most severe design conditions. For structures in Seismic 
Design Categories A, B and C, seismic force resistance would play a less significant role in the sizing of 
frame elements to control drifts, and consequently, design overstrengths for these systems would be 
somewhat lower. It seems reasonable to assume that this portion of the design overstrength for special 
moment frame structures is on the order of 2.0. 
Architectural design considerations have the potential to play a significant role in design overstrength. 
Some architectural designs will incorporate many more and larger lateral-force-resisting elements than are 
required to meet the strength and drift limitations of the code. An example of this is warehouse type 
structures, wherein the massive perimeter walls of the structure can provide very large lateral strength. 
However, even in such structures, there is typically some limiting element, such as the diaphragm, that 
prevents the design overstrength from becoming uncontrollably large. Thus, although the warehouse 
structure may have very large lateral resistance in its shear walls, typically the roof diaphragm will have a 
lateral-force-resisting capacity comparable to that specified as a minimum by the Provisions. 
Finally, the structural designer can affect the design overstrength. While some designers seek to optimize 
their structures with regard to the limitations contained in the Provisions, others will intentionally seek to 
provide greater strength and drift control than required. Typically design overstrength intentionally 
introduced by the designer will be on the order of 10 percent of the minimum required strength, but it 
may range as high as 50 to 100 percent in some cases. A factor of 1.2 should probably be presumed for 
this portion of the design overstrength to include the effects of both architectural and structural design 
2 003 Commentary, Chapter 4 
42 
overstrength. Designers who intentionally provide greater design overstrength should keep in mind that 
the O0 factors used in their designs should be adjusted accordingly. 
Material overstrength (OM) results from the fact that the design values used to proportion the elements of 
a structure are specified by the Provisions to be conservative lower bound estimates of the actual probable 
strengths of the structural materials and their effective strengths in the as-constructed structure. It is 
represented in the figure by the ratio of F2/F1, where F2 and F1 are respectively the lateral force at points 2 
and 1 on the curve. All structural materials have considerable variation in the strengths that can be 
obtained in given samples of the material from a specific grade. The design requirements typically base 
proportioning requirements on minimum specified values that are further reduced through strength 
reduction (f) factors. The actual expected strength of the as-constructed structure is significantly higher 
than this design value and should be calculated using the mean strength of the material, based on 
statistical data, by removal of the f factor from the design equation, and by providing an allowance for 
strain hardening, where significant yielding is expected to occur. Code requirements for reinforced 
masonry, concrete and steel have historically used a factor of 1.25 to account for the ratio of mean to 
specified strength and the effects of some strain hardening. Considering a typical capacity reduction 
factor on the order of 0.9, this would indicate that the material overstrength for systems constructed of 
these materials would be on the order of 1.25/0.9, or 1.4. 
System overstrength (OS) is the ratio of the ultimate lateral force the structure is capable of resisting, Fn in 
the figure, to the actual force at which first significant yield occurs, F2 in the figure. It is dependent on 
the amount of redundancy contained in the structure as well as the extent to which the designer has 
optimized the various elements that participate in lateral force resistance. For structures, with a single 
lateral-force-resisting element, such as a braced frame structure with a single bay of bracing, the system 
overstrength (OS) factor would be 1.0, because once the brace in the frame yields, the system becomes 
fully yielded. For structures that have a number of elements participating in lateral-force resistance, 
whether or not actually intended to do so, the system overstrength will be significantly larger than this, 
unless the designer has intentionally optimized the structure such that a complete side sway mechanism 
develops at the level of lateral drift at which the first actual yield occurs. 
Structural optimization is most likely to occur in structures where the actual lateral-force resistance is 
dominated by the design of elements intended to participate as part of the lateral-force-resisting system, 
and where the design of those elements is dominated by seismic loads, as opposed to gravity loads. This 
would include concentrically braced frames and eccentrically braced frames in all Seismic Design 
Categories and Special Moment Frames in Seismic Design Categories D and E. For such structures, the 
system overstrength may be taken on the order of 1.1. For dual system structures, the system 
overstrength is set by the Provisions at an approximate minimum value of 1.25. For structures where the 
number of elements that actually resist lateral forces is based on other than seismic design considerations, 
the system overstrength may be somewhat larger. In light framed residential construction, for example, 
the number of walls is controlled by architectural rather than seismic design consideration. Such 
structures may have a system overstrength on the order of 1.5. Moment frames, the design of which is 
dominated by gravity load considerations can easily have a system overstrength of 2.0 or more. This 
effect is somewhat balanced by the fact that such frames will have a lower design overstrength related to 
the requirement to increase section sizes to obtain drift control. Table C4.2-1 presents some possible 
ranges of values for the various components of overstrength for various structural systems as well as the 
overall range of values that may occur for typical structures. 
 Structural Design Criteria 
43 
Table C4.2-1 Typical Range of Overstrength for Various Systems 
Structural System 
Design 
Overstrength 
OD 
Material 
Overstrength 
OM 
System 
Overstrength 
OS 
O0 
Special moment frames (steel, concrete) 
1.5-2.5 
1.2-1.6 
1.0-1.5 
2-3.5 
Intermediate moment frames (steel, concrete) 
1.0-2.0 
1.2-1.6 
1.0-2.0 
2-3.5 
Ordinary moment frames (steel, concrete) 
1.0-1.5 
1.2-1.6 
1.5-2.5 
2-3.5 
Masonry wall frames 
1.0-2.0 
1.2-1.6 
1.0-1.5 
2-2.5 
Braced frames 
1.5-2.0 
1.2-1.6 
1.0-1.5 
1.5-2 
Reinforced bearing wall 
1.0-1.5 
1.2-1.6 
1.0-1.5 
1.5-2.5 
Reinforced infill wall 
1.0-1.5 
1.2-1.6 
1.0-1.5 
1.5-2.5 
Unreinforced bearing wall 
1.0-2.0 
0.8-2.0 
1.0-2.0 
2-3 
Unreinforced infill wall 
1.0-2.0 
0.8-2.0 
1.0-2.0 
2-3 
Dual system bracing and frame 
1.1-1.75 
1.2-1.6 
1.0-1.5 
1.5-2.5 
Light bearing wall systems 
1.0-0.5 
1.2-2.0 
1.0-2.0 
2.5-3.5 
In recognition of the fact that it is difficult to accurately estimate the amount of overstrength a structure 
will have, based solely on the type of seismic-force-resisting system that is present, in lieu of using the 
values of the overstrength coefficient O0 provided in Table 4.3-1, designers are encouraged to base the 
maximum forces used in Eq. 4.3-3 and 4.3-4 on the results of suitable nonlinear analysis of the structure. 
Such analyses should use the actual expected (rather than specified) values of material and section 
properties. Appropriate forms of such analyses could include a plastic mechanism analysis, a static 
pushover analysis, or a nonlinear response history analysis. If a plastic mechanism analysis is utilized, 
the maximum seismic force that ever could be produced in the structure, regardless of the ground motion 
experienced, is estimated. If static pushover or nonlinear response history analyses are employed, the 
forces utilized for design as the maximum force should probably be those determined for Maximum 
Considered Earthquake level ground shaking demands. 
While overstrength can be quite beneficial in permitting structures to resist actual seismic demands that 
are larger than those for which they have been specifically designed, it is not always beneficial. Some 
elements incorporated in structures behave in a brittle manner and can fail in an abrupt manner if 
substantially overloaded. The existence of structural overstrength results in a condition where such 
overloads are likely to occur, unless they are specifically accounted for in the design process. This is the 
purpose of Eq. 4.3-3 and 4.3-4. 
One case where structural overstrength should specifically be considered is in the design of column 
elements beneath discontinuous braced frames and shear walls, such as occurs at vertical in-plane and 
out-of-plane irregularities. Overstrength in the braced frames and shear walls could cause buckling 
failure of such columns with resulting structural collapse. Columns subjected to tensile loading in which 
splices are made using partial penetration groove welds, a type of joint subject to brittle fracture when 
overloaded, are another example of a case where the seismic effect with overstrength should be used. 
Other design situations that warrant the use of these equations are noted throughout the Provisions. 
Although the Provisions note the most common cases in which structural overstrength can lead to an 
undesirable failure mode, it is not possible for them to note all such conditions. Therefore, designers 
using the Provisions should be alert to conditions where the isolated independent failure of any element 
can lead to a condition of instability or collapse and should use the seismic effect with overstrength of Eq. 
2 003 Commentary, Chapter 4 
44 
4.2-3 and 4.2-4 for the design of such elements. Other conditions which may warrant such a design 
approach, although not specifically noted in the Provisions, include the design of transfer structures 
beneath discontinuous lateral-force-resisting elements and the design of diaphragm force collectors to 
shear walls and braced frames, when these are the only method of transferring force to these elements at a 
diaphragm level. 
4.3 SEISMIC-FORCE-RESISTING SYSTEM 
4.3.1 Selection and limitations. For purposes of these seismic analyses and design requirements, 
building framing systems are grouped in the structural system categories shown in Table 4.3-1. These 
categories are similar to those contained for many years in the requirements of the Uniform Building 
Code; however, a further breakdown is included for the various types of vertical components in the 
seismic-force-resisting system. In selecting a structural system, the designer is cautioned to consider 
carefully the interrelationship between continuity, toughness (including minimizing brittle behavior), and 
redundancy in the structural framing system as is subsequently discussed in this commentary. 
Specification of R factors requires considerable judgment based on knowledge of actual earthquake 
performance as well as research studies; yet, they have a major effect on building costs. The factors in 
Table 4.3-1 continue to be reviewed in light of recent research results. In the selection of the R values for 
the various systems, consideration has been given to the general observed performance of each of the 
system types during past earthquakes, the general toughness (ability to dissipate energy without serious 
degradation) of the system, and the general amount of damping present in the system when undergoing 
inelastic response. The designer is cautioned to be especially careful in detailing the more brittle types of 
systems (low Cd values). 
A bearing wall system refers to that structural support system wherein major load-carrying columns are 
omitted and the walls and/or partitions are of sufficient strength to carry the gravity loads for some 
portion of the building (including live loads, floors, roofs, and the weight of the walls themselves). The 
walls and partitions supply, in plane, lateral stiffness and stability to resist wind and earthquake loadings 
as well as any other lateral loads. In some cases, vertical trusses are employed to augment lateral 
stiffness. In general, this system has comparably lower values of R than the other systems due to the 
frequent lack of redundancy for the vertical and horizontal load support. The category designated light 
frame walls with shear panels is intended to cover wood or steel stud wall systems with finishes other 
than masonry veneers. 
A building frame system is a system in which the gravity loads are carried primarily by a frame supported 
on columns rather than by bearing walls. Some minor portions of the gravity load may be carried on 
bearing walls but the amount so carried should not represent more than a few percent of the building area. 
Lateral resistance is provided by nonbearing structural walls or braced frames. The light frame walls with 
shear panels are intended only for use with wood and steel building frames. Although there is no 
requirement to provide lateral resistance in this framing system, it is strongly recommended that some 
moment resistance be incorporated at the joints. In a structural steel frame, this could be in the form of 
top and bottom clip angles or tees at the beam- or girder-to-column connections. In reinforced concrete, 
continuity and full anchorage of longitudinal steel and stirrups over the length of beams and girders 
framing into columns would be a good design practice. With this type of interconnection, the frame 
becomes capable of providing a nominal secondary line of resistance even though the components of the 
seismic-force-resisting system are designed to carry all of the seismic force. 
A moment resisting space frame system is a system having an essentially complete space frame as in the 
building frame system. However, in this system, the primary lateral resistance is provided by moment 
resisting frames composed of columns with interacting beams or girders. Moment resisting frames may 
be either ordinary, intermediate, or special moment frames as indicated in Table 4.3-1 and limited by the 
Seismic Design Categories. 
Special moment frames must meet all the design and detailing requirements of Chapter 8, 9, 10, or 11. 
The ductility requirements for these frame systems are appropriate for all structures anticipated to 
experience large inelastic demands. For this reason, they are required in zones of high seismicity with 
 Structural Design Criteria 
45 
large anticipated ground shaking accelerations. In zones of lower seismicity, the inherent overstrength in 
typical structural designs is such that the anticipated inelastic demands are somewhat reduced, and less 
ductile systems may be safely employed. For buildings in which these special design and detailing 
requirements are not used, lower R values are specified indicating that ordinary framing systems do not 
possess as much toughness and that less reduction from the elastic response can be tolerated. 
Requirements for composite steel-concrete systems were first introduced in the 1994 Edition. The R, O0, 
and Cd values for the composite systems in Table 4.3-1 are similar to those for comparable systems of 
structural steel and reinforced concrete. The values shown in Table 4.3-1 are only allowed when the 
design and detailing requirements for composite structures in Chapter 10 are followed. 
Inverted pendulum structures are singled out for special consideration because of their unique 
characteristics. These structures have little redundancy and overstrength and concentrate inelastic 
behavior at their bases. As a result, they have substantially less energy dissipation capacity than other 
systems. A number of buildings incorporating this system experienced very severe damage, and in some 
cases, collapse, in the 1994 Northridge earthquake. 
4.3.1.1 Dual system. A dual system consists of a three-dimensional space frame made up of columns 
and beams that provide primary support for the gravity loads. Primary lateral resistance is supplied by 
structural nonbearing walls or bracing; the frame is provided with a redundant lateral-force-resisting 
system that is a moment frame complying with the requirements of Chapters 8, 9, 10, or 11. The moment 
frame is required to be capable of resisting at least 25 percent of the specified seismic force; this 
percentage is based on the judgment of the writers. Normally the moment frame would be a part of the 
basic space frame. The walls or bracing acting together with the moment frame must be capable of 
resisting all of the design seismic force. The following analyses are required for dual systems: 
1. The frame and shear walls or braced frames must resist the prescribed lateral seismic force in 
accordance with their relative rigidities considering fully the interaction of the walls or braced frames 
and the moment frames as a single system. This analysis must be made in accordance with the 
principles of structural mechanics considering the relative rigidities of the elements and torsion in the 
system. Deformations imposed upon members of the moment frame by their interaction with the 
shear walls or braced frames must be considered in this analysis. 
2. The moment frame must be designed to have a capacity to resist at least 25 percent of the total 
required lateral seismic force including torsional effects. 
4.3.1.2 Combinations of framing systems. For those cases where combinations of structural systems 
are employed, the designer must use judgment in selecting appropriate R, O0, and Cd values. The intent of 
Sec. 4.3.1.2.1 is to prohibit support of one system by another possessing characteristics that result in a 
lower base shear factor. The entire system should be designed for the higher seismic shear as the 
provision stipulates. The exception is included to permit the use of such systems as a braced frame 
penthouse on a moment frame building in which the mass of the penthouse does not represent a 
significant portion of the total building and, thus, would not materially affect the overall response to 
earthquake motions. 
Sec. 4.3.1.2.2 pertains to details and is included to help ensure that the more ductile details inherent with 
the design for the higher R value system will be employed throughout. The intent is that details common 
to both systems be designed to remain functional throughout the response in order to preserve the 
integrity of the seismic-force-resisting system. 
4.3.1.3 - 4.3.1.6 Seismic Design Categories. General framing system requirements for the building 
Seismic Design Categories are given in these sections. The corresponding design and detailing 
requirements are given in Sec. 4.6 and Chapters 8 through 14. There are no restrictions on the selection 
of structural systems in Seismic Design Category A. Table 4.3-1 indicates the systems permitted in all 
other Seismic Design Categories. 
2 003 Commentary, Chapter 4 
46 
4.3.1.4 Seismic Design Category D. Sec. 4.3.1.4 covers Seismic Design Category D, which compares 
roughly to California design practice for normal buildings away from major faults. In keeping with the 
philosophy of present codes for zones of high seismic risk, these requirements continue limitations on the 
use of certain types of structures over 160 ft (49 m) in height but with some changes. Although it is 
agreed that the lack of reliable data on the behavior of high-rise buildings whose structural systems 
involve shear walls and/or braced frames makes it convenient at present to establish some limits, the 
values of 160 ft (49 m) and 240 ft (73 m) introduced in these requirements are arbitrary. Considerable 
disagreement exists regarding the adequacy of these values, and it is intended that these limitations be the 
subject of further study. 
According to these requirements require that buildings in Category D over 160 ft (49 m) in height must 
have one of the following seismic-force-resisting systems: 
1. A moment resisting frame system with special moment frames capable of resisting the total 
prescribed seismic force. This requirement is the same as present SEAOC and UBC 
recommendations. 
2. A dual system as defined in this chapter, wherein the prescribed forces are resisted by the entire 
system and the special moment frame is designed to resist at least 25 percent of the prescribed seismic 
force. This requirement is also similar to SEAOC and UBC recommendations. The purpose of the 25 
percent frame is to provide a secondary defense system with higher degrees of redundancy and 
ductility in order to improve the ability of the building to support the service loads (or at least the 
effect of gravity loads) after strong earthquake shaking. It should be noted that SEAOC and UBC 
requirements prior to 1987 required that shear walls or braced frames be able to resist the total 
required seismic lateral forces independently of the special moment frame. The Provisions require 
only that the true interaction behavior of the frame-shear wall (or braced frame) system be 
considered. If the analysis of the interacting behavior is based only on the vertical distribution of 
seismic lateral forces determined using the equivalent lateral force procedure of Sec. 5.2, the 
interpretation of the results of this analysis for designing the shear walls or braced frame should 
recognize the effects of higher modes of vibration. The internal forces that can be developed in the 
shear walls in the upper stories can be more severe than those obtained from the ELF procedure. 
3. The use of a shear wall (or braced frame) system of cast-in-place concrete or structural steel up to a 
height of 240 ft (73 m) is permitted only if braced frames or shear walls in any plane do not resist 
more than 60 percent of the seismic design force including torsional effects and the configuration of 
the lateral-force-resisting system is such that torsional effects result in less than a 20 percent 
contribution to the strength demand on the walls or frames. The intent is that each of these shear 
walls or braced frames be in a different plane and that the four or more planes required be spaced 
adequately throughout the plan or on the perimeter of the building in such a way that the premature 
failure of one of the single walls or frames will not lead to excessive inelastic torsion. 
Although a structural system with lateral force resistance concentrated in the interior core (Figure C4.3-1 
is acceptable according to the Provisions, it is highly recommended that use of such a system be avoided, 
particularly for taller buildings. The intent is to replace it by the system with lateral force resistance 
distributed across the entire building (Figure C4.3-2). The latter system is believed to be more suitable in 
view of the lack of reliable data regarding the behavior of tall buildings having structural systems based 
on central cores formed by coupled shear walls or slender braced frames. 
 Structural Design Criteria 
47 
Figure C4.3-1 Arrangement of shear walls and 
braced frames  not recommended. Note that the 
heavy lines indicate shear walls and/or braced 
frames. 
Figure C4.3-2 Arrangement of shear walls and 
braced frames  recommended. Note that the 
heavy lines indicate shear walls and/or braced 
frames. 
4.3.1.4.2 Interaction effects. This section relates to the interaction of elements of the seismic-forceresisting 
system with elements that are not part of this system. A classic example of such interaction is 
the behavior of infill masonry walls used as architectural elements in a building provided with a seismicforce-
resisting system composed of moment resisting frames. Although the masonry walls are not 
intended to resist seismic forces, at low levels of deformation they will be substantially more rigid than 
the moment resisting frames and will participate in lateral force resistance. A common effect of such 
walls is that they can create shear-critical conditions in the columns they abut by reducing the effective 
flexural height of these columns to the height of the openings in the walls. If these walls are neither 
uniformly distributed throughout the structure nor effectively isolated from participation in lateral force 
resistance, they can also create torsional irregularities and soft story irregularities in structures that would 
otherwise have regular configuration. 
Infill walls are not the only elements not included in seismic-force-resisting systems that can affect a 
structure=s seismic behavior. For example, in parking garage structures, the ramps between levels can act 
as effective bracing elements and resist a large portion of the seismically induced forces. They can induce 
large thrusts in the diaphragms where they connect, as well as large vertical forces on the adjacent 
columns and beams. In addition, if not symmetrically placed in the structure they can induce torsional 
irregularities. This section requires consideration of these potential effects. 
4.3.1.6 Seismic Design Category F. Sec. 4.3.1.6 covers Category F, which is restricted to essential 
facilities on sites located within a few kilometers of major active faults. Because of the necessity for 
reducing risk (particularly in terms of providing life safety or maintaining function by minimizing 
damage to nonstructural building elements, contents, equipment, and utilities), the height limitations for 
Category F are reduced. Again, the limits100 ft (30 m) and 160 ft (49 m)are arbitrary and require 
further study. The developers of these requirements believe that, at present, it is advisable to establish 
these limits, but the importance of having more stringent requirements for detailing the seismic-forceresisting 
system as well as the nonstructural components of the building must be stressed. Such 
requirements are specified in Sec. 4.6 and Chapters 8 through 12. 
4.3.2 Configuration. The configuration of a structure can significantly affect its performance during a 
strong earthquake that produces the ground motion contemplated in the Provisions. Configuration can be 
divided into two aspects: plan configuration and vertical configuration. The Provisions were basically 
derived for buildings having regular configurations. Past earthquakes have repeatedly shown that 
buildings having irregular configurations suffer greater damage than buildings having regular 
configurations. This situation prevails even with good design and construction. There are several reasons 
2 003 Commentary, Chapter 4 
48 
for this poor behavior of irregular structures. In a regular structure, inelastic demands produced by strong 
ground shaking tend to be well distributed throughout the structure, resulting in a dispersion of energy 
dissipation and damage. However, in irregular structures, inelastic behavior can concentrate in the zone 
of irregularity, resulting in rapid failure of structural elements in these areas. In addition, some 
irregularities introduce unanticipated stresses into the structure which designers frequently overlook when 
detailing the structural system. Finally, the elastic analysis methods typically employed in the design of 
structures often cannot predict the distribution of earthquake demands in an irregular structure very well, 
leading to inadequate design in the zones of irregularity. For these reasons, these requirements are 
designed to encourage that buildings be designed to have regular configurations and to prohibit gross 
irregularity in buildings located on sites close to major active faults, where very strong ground motion and 
extreme inelastic demands can be experienced. 
4.3.2.2 Plan irregularity. Sec. 4.3.2.2 indicates, by reference to Table 4.3-2, under what circumstances 
a building must be designated as having a plan irregularity for the purposes of the Provisions. A building 
may have a symmetrical geometric shape without re-entrant corners or wings but still be classified as 
irregular in plan because of distribution of mass or vertical, seismic-force-resisting elements. Torsional 
effects in earthquakes can occur even when the static centers of mass and resistance coincide. For 
example, ground motion waves acting with a skew with respect to the building axis can cause torsion. 
Cracking or yielding in a nonsymmetrical fashion also can cause torsion. These effects also can magnify 
the torsion due to eccentricity between the static centers. For this reason, buildings having an eccentricity 
between the static center of mass and the static center of resistance in excess of 10 percent of the building 
dimension perpendicular to the direction of the seismic force should be classified as irregular. The 
vertical resisting components may be arranged so that the static centers of mass and resistance are within 
the limitations given above and still be unsymmetrically arranged so that the prescribed torsional forces 
would be unequally distributed to the various components. In the 1997 Provisions, torsional irregularities 
were subdivided into two categories, with a category of extreme irregularity having been created. 
Extreme torsional irregularities are prohibited for structures located very close to major active faults and 
should be avoided, when possible, in all structures. 
There is a second type of distribution of vertical, resisting components that, while not being classified as 
irregular, does not perform well in strong earthquakes. This arrangement is termed a core-type building 
with the vertical components of the seismic-force-resisting system concentrated near the center of the 
building. Better performance has been observed when the vertical components are distributed near the 
perimeter of the building. In recognition of the problems leading to torsional instability, a torsional 
amplification factor is introduced in Sec. 5.2.4.3. 
A building having a regular configuration can be square, rectangular, or circular. A square or rectangular 
building with minor re-entrant corners would still be considered regular but large re-entrant corners 
creating a crucifix form would be classified as an irregular configuration. The response of the wings of 
this type of building is generally different from the response of the building as a whole, and this produces 
higher local forces than would be determined by application of the Provisions without modification. 
Other plan configurations such as H-shapes that have a geometrical symmetry also would be classified as 
irregular because of the response of the wings. 
Significant differences in stiffness between portions of a diaphragm at a level are classified as 
irregularities since they may cause a change in the distribution of seismic forces to the vertical 
components and create torsional forces not accounted for in the normal distribution considered for a 
regular building. Examples of plan irregularities are illustrated in Figure C4.3-3. 
 Structural Design Criteria 
49 
Figure 4.3-3 Building plan irregularities 
Where there are discontinuities in the path of lateral force resistance, the structure can no longer be 
considered to be regular. The most critical of the discontinuities to be considered is the out-of-plane 
offset of vertical elements of the seismic-force-resisting elements. Such offsets impose vertical and 
lateral load effects on horizontal elements that are, at the least, difficult to provide for adequately. 
Where vertical elements of the lateral-force-resisting system are not parallel to or symmetric about major 
orthogonal axes, the static lateral force procedures of the Provisions cannot be applied as given and, thus, 
the structure must be considered to be irregular. 
4.3.2.3 Vertical irregularity. Sec. 4.3.2.3 indicates, by reference to Table 4.3-3, under what 
circumstances a structure must be considered to have a vertical irregularity. Vertical configuration 
irregularities affect the responses at the various levels and induce loads at these levels that are 
significantly different from the distribution assumed in the equivalent lateral force procedure given in 
Sec. 5.2. 
A moment resisting frame building might be classified as having a vertical irregularity if one story were 
much taller than the adjoining stories and the design did not compensate for the resulting decrease in 
stiffness that would normally occur. Examples of vertical irregularities are illustrated in Figure C4.3-4. 
2 003 Commentary, Chapter 4 
50 
Figure C4.3-4 Building elevation irregularities 
A building would be classified as irregular if the ratio of mass to stiffness in adjoining stories differs 
significantly. This might occur when a heavy mass, such as a swimming pool, is placed at one level. 
Note that the exception in the Provisions provides a comparative stiffness ratio between stories to exempt 
structures from being designated as having a vertical irregularity of the types specified. 
One type of vertical irregularity is created by unsymmetrical geometry with respect to the vertical axis of 
the building. The building may have a geometry that is symmetrical about the vertical axis and still be 
classified as irregular because of significant horizontal offsets in the vertical elements of the lateral-forceresisting 
system at one or more levels. An offset is considered to be significant if the ratio of the larger 
dimension to the smaller dimension is more than 130 percent. The building also would be considered 
irregular if the smaller dimension were below the larger dimension, thereby creating an inverted pyramid 
effect. 
Weak story irregularities occur whenever the strength of a story to resist lateral demands is significantly 
less than that of the story above. This is because buildings with this configuration tend to develop all of 
their inelastic behavior at the weak story. This can result in a significant change in the deformation 
pattern of the building, with most earthquake induced displacement occurring within the weak story. This 
can result in extensive damage within the weak story and even instability and collapse. Note that an 
exception has been provided in Sec. 4.6.1.6 where there is considerable overstrength of the weak story. 
In the 1997 Provisions, the soft story irregularity was subdivided into two categories with an extreme soft 
story category being created. Like weak stories, soft stories can lead to instability and collapse. 
Buildings with extreme soft stories are now prohibited on sites located very close to major active faults. 
4.3.3 Redundancy. The 1997 Provisions introduced specific requirements intended to quantify the 
importance of redundancy. Many parts of the Provisions, particularly the response modification 
 Structural Design Criteria 
51 
coefficients, R, were originally developed assuming that structures possess varying levels of redundancy 
that heretofore were undefined. Commentary Sec. 4.2.1 recommends that lower R values be used for nonredundant 
systems, but does not provide guidance on how to select and justify appropriate reductions. As 
a result, many non-redundant structures have been designed in the past using values of R that were 
intended for use in designing structures with higher levels of redundancy. For example, current R values 
for special moment resisting frames were initially established in the 1970s based on the then widespread 
use of complete or nearly complete frame systems in which all beam-column connections were designed 
to participate in the lateral-force-resisting system. High R values were justified by the large number of 
potential hinges that could form in such redundant systems, and the beneficial effects of progressive yield 
hinge formation described in Sec. C4.2.1. However, in recent years, economic pressures have encouraged 
the now prevalent use of much less redundant special moment frames with relatively few bays of moment 
resisting framing supporting large floor and roof areas. Similar observations have been made of other 
types of construction as well. Modern concrete and masonry shear wall buildings, for example, have 
many fewer walls than were once commonly provided in such buildings. 
In order to quantify the effects of redundancy, the 1997 Provisions introduced the concept of a 
redundancy factor, ., that is applied to the design earthquake loads in the seismic load effect equations of 
Sec. 4.2.2.1, for structures in Seismic Design Categories D, E, and F. The value of the reliability factor . 
varies from 1 to 1.5. In effect this reduces the R values for less redundant structures and should provide 
greater economic incentive for the design of structures with well distributed lateral-force-resisting 
systems. The formulation for the equation from which . is derived is similar to that developed by 
SEAOC for inclusion in the 1997 edition of the Uniform Building Code. It bases the value of . on the 
floor area of the building and the parameter r which relates to the amount of the building=s design 
lateral force carried by any single element. 
There are many other considerations than just floor area and element/story shear ratios that should be 
considered in quantifying redundancy. Conceptually, element demand/capacity ratios, types of 
mechanisms which may form, individual characteristics of building systems and materials, building 
height, number of stories, irregularity, torsional resistance, chord and collector length, diaphragm spans, 
number of lines of resistance, and number of elements per line are all important and will intrinsically 
influence the level of redundancy in systems and their reliability. 
The SEAOC proposed code change to the 1997 UBC recommends addressing redundancy in irregular 
buildings by evaluating the ratio of element shear to design story shear, r only in the lower two-thirds 
of the height. However, in response to failures of buildings that have occurred at and above mid-heights, 
the writers of the Provisions chose to base the . factor on the worst r for the least redundant story. The 
resulting factor is then applied throughout the height of the building. 
The Applied Technology Council, in its ATC 19 report suggests that future redundancy factors be based 
on reliability theory. For example, if the number of hinges in a moment frame required to achieve a 
minimally redundant system were established, a redundancy factor for less redundant systems could be 
based on the relationship of the number of hinges actually provided to those required for minimally 
redundant systems. ATC suggests that similar relationships could be developed for shear wall systems 
using reliability theory. However, much work yet remains to be completed before such approaches will 
be ready for adoption into the Provisions. 
The Provisions limit special moment resisting frames to configurations that provide maximum . values of 
1.25 and 1.1, respectively, in Seismic Design Categories D, and E or F, to compensate for the strength 
based factor in what are typically drift-controlled systems. Other seismic-force-resisting systems that are 
not typically drift controlled may be proportioned to exceed the maximum . factor of 1.5; however, it is 
not recommended that this be done. 
2 003 Commentary, Chapter 4 
52 
4.4 STRUCTURAL ANALYSIS 
4.4.1 Procedure selection. Many of the standard procedures for the analysis of forces and deformations 
in structures subjected to earthquake ground motion are listed below in order of increasing rigor and 
expected accuracy: 
1. Equivalent lateral force procedure (Sec. 5.2). 
2. Response spectrum (modal analysis) procedure (Sec. 5.3). 
3. Linear response history procedure (Sec. 5.4). 
4. Nonlinear static procedure, involving incremental application of a pattern of lateral forces and 
adjustment of the structural model to account for progressive yielding under load application (pushover 
analysis) (Appendix to Chapter 5). 
5. Nonlinear response history procedure involving step-by-step integration of the coupled equations of 
motion (Sec. 5.5). 
Each procedure becomes more rigorous if effects of soil-structure interaction are considered, either as 
presented in Sec. 5.6 or through a more complete analysis of this interaction, as appropriate. Every 
procedure improves in rigor if combined with use of results from experimental research (not described in 
these Provisions). 
4.4.2 Application of loading. Earthquake forces act in both principal directions of the building 
simultaneously, but the earthquake effects in the two principal directions are unlikely to reach their 
maxima simultaneously. This section provides a reasonable and adequate method for combining them. It 
requires that structural elements be designed for 100 percent of the effects of seismic forces in one 
principal direction combined with 30 percent of the effects of seismic forces in the orthogonal direction. 
The following combinations of effects of gravity loads, effects of seismic forces in the x-direction, and 
effects of seismic forces in the y-direction (orthogonal to x-direction) thus pertain: 
gravity " 100% of x-direction " 30% of y-direction 
gravity " 30% of x-direction " 100% of y-direction 
The combination and signs (plus or minus) requiring the greater member strength are used for each 
member. Orthogonal effects are slight on beams, girders, slabs, and other horizontal elements that are 
essentially one-directional in their behavior, but they may be significant in columns or other vertical 
members that participate in resisting earthquake forces in both principal directions of the building. For 
two-way slabs, orthogonal effects at slab-to-column connections can be neglected provided the moment 
transferred in the minor direction does not exceed 30 percent of that transferred in the orthogonal 
direction and there is adequate reinforcement within lines one and one-half times the slab thickness either 
side of the column to transfer all the minor direction moment. 
4.5 DEFORMATION REQUIREMENTS 
4.5.1 Deflection and drift limits. This section provides procedures for the limitation of story drift. The 
term drift has two connotations: 
1. Story drift is the maximum lateral displacement within a story (i.e., the displacement of one floor 
relative to the floor below caused by the effects of seismic loads). 
2. The lateral displacement or deflection due to design forces is the absolute displacement of any point 
in the structure relative to the base. This is not story drift and is not to be used for drift control or 
stability considerations since it may give a false impression of the effects in critical stories. However, 
it is important when considering seismic separation requirements. 
There are many reasons for controlling drift; one is to control member inelastic strain. Although use of 
drift limitations is an imprecise and highly variable way of controlling strain, this is balanced by the 
current state of knowledge of what the strain limitations should be. 
 Structural Design Criteria 
53 
Stability considerations dictate that flexibility be controlled. The stability of members under elastic and 
inelastic deformation caused by earthquakes is a direct function of both axial loading and bending of 
members. A stability problem is resolved by limiting the drift on the vertical-load-carrying elements and 
the resulting secondary moment from this axial load and deflection (frequently called the P-delta effect). 
Under small lateral deformations, secondary stresses are normally within tolerable limits. However, 
larger deformations with heavy vertical loads can lead to significant secondary moments from the P-delta 
effects in the design. The drift limits indirectly provide upper bounds for these effects. 
Buildings subjected to earthquakes need drift control to restrict damage to partitions, shaft and stair 
enclosures, glass, and other fragile nonstructural elements and, more importantly, to minimize differential 
movement demands on the seismic safety elements. Since general damage control for economic reasons 
is not a goal of this document and since the state of the art is not well developed in this area, the drift 
limits have been established without regard to considerations such as present worth of future repairs 
versus additional structural costs to limit drift. These are matters for building owners and designers to 
examine. To the extent that life might be excessively threatened, general damage to nonstructural and 
seismic-safety elements is a drift limit consideration. 
The design story drift limits of Table 4.5-1 reflect consensus judgment taking into account the goals of 
drift control outlined above. In terms of life safety and damage control objectives, the drift limits should 
yield a substantial, though not absolute, measure of safety for well detailed and constructed brittle 
elements and provide tolerable limits wherein the seismic safety elements can successfully perform, 
provided they are designed and constructed in accordance with these Provisions. 
To provide a higher performance standard, the drift limit for the essential facilities of Seismic Use 
Group III is more stringent than the limit for Groups I and II except for masonry shear wall buildings. 
The drift limits for low-rise structures are relaxed somewhat provided the interior walls, partitions, 
ceilings, and exterior wall systems have been designed to accommodate story drifts. The type of steel 
building envisioned by the exception to the table would be similar to a prefabricated steel structure with 
metal skin. When the more liberal drift limits are used, it is recommended that special requirements be 
provided for the seismic safety elements to accommodate the drift. 
It should be emphasized that the drift limits, .a, of Table 4.5-1 are story drifts and, therefore, are 
applicable to each story (that is, they must not be exceeded in any story even though the drift in other 
stories may be well below the limit). The limit, .a is to be compared to the design story drift as 
determined by Sec. 5.2.6.1. 
Stress or strength limitations imposed by design level forces occasionally may provide adequate drift 
control. However, it is expected that the design of moment resisting frames, especially steel building 
frames, and the design of tall, narrow shear wall or braced frame buildings will be governed at least in 
part by drift considerations. In areas having large design spectral response accelerations, SDS and SD1, it is 
expected that seismic drift considerations will predominate for buildings of medium height. In areas 
having a low design spectral response accelerations and for very tall buildings in areas with large design 
spectral response accelerations, wind considerations generally will control, at least in the lower stories. 
Due to probable first mode drift contributions, the Sec. 5.2 ELF procedure may be too conservative for 
drift design of very tall moment-frame buildings. It is suggested for these buildings, where the first mode 
would be responding in the constant displacement region of a response spectra (where displacements 
would be essentially independent of stiffness), that the response spectrum procedure of Sec. 5.3 be used 
for design even when not required by Sec. 4.4.1. 
Building separations and seismic joints are separations between two adjoining buildings or parts of the 
same building, with or without frangible closures, for the purpose of permitting the adjoining buildings or 
parts to respond independently to earthquake ground motion. Unless all portions of the structure have 
been designed and constructed to act as a unit, they must be separated by seismic joints. For irregular 
structures that cannot be expected to act reliably as a unit, seismic joints should be utilized to separate the 
building into units whose independent response to earthquake ground motion can be predicted. 
2 003 Commentary, Chapter 4 
54 
Although the Provisions do not give precise formulations for the separations, it is required that the 
distance be sufficient to avoid damaging contact under total deflection in order to avoid interference 
and possible destructive hammering between buildings. It is recommended that the distance be equal to 
the total of the lateral deflections of the two units assumed deflecting toward each other (this involves 
increasing separations with height). If the effects of hammering can be shown not to be detrimental, these 
distances can be reduced. For very rigid shear wall structures with rigid diaphragms whose lateral 
deflections cannot be reasonably estimated, it is suggested that older code requirements for structural 
separations of at least 1 in. (25 mm) plus 1/2 in. (13 mm) for each 10 ft (3 m) of height above 20 ft (6 m) 
be followed. 
4.5.3 Seismic Design Categories D, E, and F. The purpose of this section is to require that the seismicforce-
resisting system provide adequate deformation control to protect elements of the structure that are 
not part of the seismic-force-resisting system. In regions of high seismicity, it is relatively common to 
apply ductile detailing requirements to elements which are intended to resist seismic forces but to neglect 
such practices in nonstructural elements or elements intended to resist only gravity forces. The fact that 
many elements of the structure are not intended to resist seismic forces and are not detailed for such 
resistance does not prevent them from actually providing this resistance and becoming severely damaged 
as a result. 
The 1994 Northridge earthquake provided several examples where this was a cause of failure. In a 
preliminary reconnaissance report of that earthquake (EERI, 1994) it was stated: Of much significance 
is the observation that six of the seven partial collapses (in modern precast concrete parking structures) 
seem to have been precipitated by damage to the gravity load system. Possibly, the combination of large 
lateral deformation and vertical load caused crushing in poorly confined columns that were not detailed to 
be part of the lateral load resisting system. The report also noted that: Punching shear failures were 
observed in some structures at slab-to-column connections such as at the Four Seasons building in 
Sherman Oaks. The primary lateral load resisting system was a perimeter ductile frame that performed 
quite well. However, the interior slab-column system was incapable of undergoing the same lateral 
deflections and experienced punching failures. 
In response to a preponderance of evidence, SEAOC successfully submitted a change to the Uniform 
Building Code in 1994 to clarify and strengthen the existing requirements intended to require deformation 
compatibility. The statement in support of that code change included the following reasons: 
Deformation compatibility requirements have largely been ignored by the design community. In the 
1994 Northridge earthquake, deformation-induced damage to elements which were not part of the lateralforce-
resisting system resulted in structural collapse. Damage to elements of the lateral-framing system, 
whose behavior was affected by adjoining rigid elements, was also observed. This has demonstrated a 
need for stronger and clearer requirements. The proposed changes attempt to emphasize the need for 
specific design and detailing of elements not part of the lateral system to accommodate expected seismic 
deformation. . . . 
Language introduced in the 1997 Provisions was largely based on SEAOCs successful 1995 change to 
the Uniform Building Code. Rather than implicitly relying on designers to assume appropriate levels of 
stiffness, the language in Sec. 4.5.3 explicitly requires that the stiffening effects of adjoining rigid 
structural and nonstructural elements shall be considered and a rational value of member and restraint 
stiffness shall be used for the design of components that are not part of the lateral-force-resisting system. 
This was intended to keep designers from neglecting the potentially adverse stiffening effects that such 
components can have on structures. This section also includes a requirement to address shears that can be 
induced in structural components that are not part of the lateral-force-resisting system since sudden shear 
failures have been catastrophic in past earthquakes. 
The exception in Sec. 4.5.3 is intended to encourage the use of intermediate or special detailing in beams 
and columns that are not part of the lateral-force-resisting system. In return for better detailing, such 
beams and columns are permitted to be designed to resist moments and shears from unamplified 
deflections. This reflects observations and experimental evidence that well-detailed components can 
 Structural Design Criteria 
55 
accommodate large drifts by responding inelastically without losing significant vertical load carrying 
capacity. 
4.6 DESIGN AND DETAILING REQUIREMENTS 
The design and detailing requirements for components of the seismic-force-resisting system are stated in 
this section. The requirements of this section are spelled out in considerable detail. The major reasons 
for this are presented below. 
The provision of detailed design ground motions and requirements for analysis of the structure do not by 
themselves make a building earthquake resistant. Additional design requirements are necessary to 
provide a consistent degree of earthquake resistance in buildings. The more severe the expected seismic 
ground motions, the more stringent these additional design requirements should be. Not all of the 
necessary design requirements are expressed in codes, and although experienced seismic design engineers 
account for them, engineers lacking experience in the design and construction of earthquake-resistant 
structures often overlook them. Considerable uncertainties exist regarding: 
1. The actual dynamic characteristics of future earthquake motions expected at a building site; 
2. The soil-structure-foundation interaction; 
3. The actual response of buildings when subjected to seismic motions at their foundations; and 
4. The mechanical characteristics of the different structural materials, particularly when they undergo 
significant cyclic straining in the inelastic range that can lead to severe reversals of strains. 
It should be noted that the overall inelastic response of a structure is very sensitive to the inelastic 
behavior of its critical regions, and this behavior is influenced, in turn, by the detailing of these regions. 
Although it is possible to counteract the consequences of these uncertainties by increasing the level of 
design forces, it is considered more feasible to provide a building system with the largest energy 
dissipation consistent with the maximum tolerable deformations of nonstructural components and 
equipment. This energy dissipation capacity, which is usually denoted simplistically as ductility, is 
extremely sensitive to the detailing. Therefore, in order to achieve such a large energy dissipation 
capacity, it is essential that stringent design requirements be used for detailing the structural as well as the 
nonstructural components and their connections or separations. Furthermore, it is necessary to have good 
quality control of materials and competent inspection. The importance of these factors has been clearly 
demonstrated by the building damage observed after both moderate and severe earthquakes. 
It should be kept in mind that a buildings response to seismic ground motion most often does not reflect 
the designers or analysts original conception or modeling of the structure on paper. What is reflected is 
the manner in which the building was constructed in the field. These requirements emphasize the 
importance of detailing and recognize that the detailing requirements should be related to the expected 
earthquake intensities and the importance of the buildings function and/or the density and type of 
occupancy. The greater the expected intensity of earthquake ground-shaking and the more important the 
building function or the greater the number of occupants in the building, the more stringent the design and 
detailing requirements should be. In defining these requirements, the Provisions uses the concept of 
Seismic Design Categories (Tables 1.4-1 and 1.4-2), which relate to the design ground motion severities, 
given by the spectral response acceleration coefficients SDS and SD1 (Chapter 3) and the Seismic Use 
Group (Sec. 1.2). 
4.6.1 Seismic Design Category B. Category B and Category C buildings will be constructed in the 
largest portion of the United States. Earthquake-resistant requirements are increased appreciably over 
Category A requirements, but they still are quite simple compared to present requirements in areas of high 
seismicity. 
The Category B requirements specifically recognize the need to design diaphragms, provide collector 
bars, and provide reinforcing around openings. There requirements may seem elementary and obvious 
but, because they are not specifically covered in many codes, some engineers totally neglect them. 
2 003 Commentary, Chapter 4 
56 
4.6.1.1 Connections. The analysis of a structure and the provision of a design ground motion alone do 
not make a structure earthquake resistant; additional design requirements are necessary to provide 
adequate earthquake resistance in structures. Experienced seismic designers normally fill these 
requirements, but because some were not formally specified, they often are overlooked by inexperienced 
engineers. 
Probably the most important single attribute of an earthquake-resistant structure is that it is tied together 
to act as a unit. This attribute is important not only in earthquake-resistant design, but also is 
indispensable in resisting high winds, floods, explosion, progressive failure, and even such ordinary 
hazards as foundation settlement. This section requires that all parts of the building (or unit if there are 
separation joints) be so tied together that any part of the structure is tied to the rest to resist a force of 
0.133SDS (but not less than 0.05) times the weight of the smaller. In addition, beams must be tied to their 
supports or columns and columns to footings for a minimum of 5 percent of the dead and live load 
reaction. 
Certain connections of buildings with plan irregularities must be designed for higher forces than 
calculated due to the simplifying assumptions used in the analysis by Sec. 5.2, 5.3, and 5.4 (see Sec. 
4.6.3.2). 
4.6.1.2 Anchorage of concrete or masonry walls. One of the major hazards from buildings during an 
earthquake is the pulling away of heavy masonry or concrete walls from floors or roofs. Although 
requirements for the anchorage to prevent this separation are common in highly seismic areas, they have 
been minimal or nonexistent in most other parts of the country. This section requires that anchorage be 
provided in any locality to the extent of 400SDS pounds per linear foot (plf) or 5,840 times SDS (N/m). 
This requirement alone may not provide complete earthquake-resistant design, but observations of 
earthquake damage indicate that it can greatly increase the earthquake resistance of buildings and reduce 
hazards in those localities where earthquakes may occur but are rarely damaging. 
4.6.1.3 Bearing walls. A minimum anchorage of bearing walls to diaphragms or other resisting elements 
is specified. To ensure that the walls and supporting framing system interact properly, it is required that 
the interconnection of dependent wall elements and connections to the framing system have sufficient 
ductility or rotational capacity, or strength, to stay as a unit. Large shrinkage or settlement cracks can 
significantly affect the desired interaction. 
4.6.1.5 Inverted pendulum-type structures. Inverted pendulum-type structures have a large portion of 
their mass concentrated near the top and, thus, have essentially one degree of freedom in horizontal 
translation. Often the structures are T-shaped with a single column supporting a beam or slab at the top. 
For such a structure, the lateral motion is accompanied by rotation of the horizontal element of the T 
due to rotation at the top of the column, resulting in vertical accelerations acting in opposite directions on 
the overhangs of the structure. Dynamic response amplifies this rotation; hence, a bending moment 
would be induced at the top of the column even though the procedures of Sec. 5.2 would not so indicate. 
A simple provision to compensate for this is specified in this section. The bending moments due to the 
lateral force are first calculated for the base of the column according to the requirements of Sec. 5.2. 
One-half of the calculated bending moment at the base is applied at the top and the moments along the 
column are varied from 1.5 M at the base to 0.5 M at the top. The addition of one-half the moment 
calculated at the base in accordance with Sec. 5.2 is based on analyses of inverted pendulums covering a 
wide range of practical conditions. 
4.6.1.8 Collector elements. Many buildings have shear walls or other bracing elements that are not 
uniformly spaced around the diaphragms. Such conditions require that collector or drag members be 
provided. A simple illustration is shown in Figure C4.6-1. 
 Structural Design Criteria 
57 
Figure 4.6-1 Collector element used to (a) transfer shears and (b) 
transfer drag forces from diaphragm to shear wall 
Consider a building as shown in the plan with four short shear walls at the corners arranged as shown. 
For north-south earthquake forces, the diaphragm shears on Line AB are uniformly distributed between A 
and B if the chord reinforcing is assumed to act on Lines BC and AD. However, wall A is quite short so 
reinforcing steel is required to collect these shears and transfer them to the wall. If Wall A is a quarter of 
the length of AB, the steel must carry, as a minimum, three-fourths of the total shear on Line AB. The 
same principle is true for the other walls. In Figure C4.6-1 reinforcing is required to collect the shears or 
drag the forces from the diaphragm into the shear wall. Similar collector elements are needed for most 
shear walls and for some frames. 
4.6.1.9 Diaphragms. Diaphragms are deep beams or trusses that distribute the lateral loads from their 
origin to the components where such forces are resisted. Therefore, diaphragms are subject to shears, 
bending moments, direct stresses (truss member, collector elements), and deformations. The 
deformations must be minimized in some cases because they could overstress the walls to which the 
diaphragms are connected. The amount of deflection permitted in the diaphragm must be related to the 
ability of the walls to deflect (normal to the direction of force application) without failure. 
A detail commonly overlooked by many engineers is the requirement to tie the diaphragm together so that 
it acts as a unit. Wall anchorages tend to tear off the edges of the diaphragm; thus, the ties must be 
extended into the diaphragm so as to develop adequate anchorage. During the San Fernando earthquake, 
seismic forces from the walls caused separations in roof diaphragms 20 or more feet (6 m) from the edge 
in several industrial buildings. 
Where openings occur in shear walls or diaphragms, temperature trim bars alone do not provide 
adequate reinforcement. The chord stresses must be provided for and the chords anchored to develop the 
chord stresses by embedment. The embedment must be sufficient to take the reactions without 
2 003 Commentary, Chapter 4 
58 
overstressing the material in any respect. Since the design basis depends on an elastic analysis, the 
internal force system should be compatible with both static and the elastic deformations. 
4.6.1.10 Anchorage of nonstructural systems. Anchorage of nonstructural systems and components of 
buildings is required as indicated in Chapter 6. 
4.6.2 Seismic Design Category C. The material requirements in Chapters 8 through 12 for Category C 
are somewhat more restrictive than those for Categories A and B. Also, a nominal interconnection 
between pile caps and caissons is required. 
4.6.3 Seismic Design Categories D, E and F. Category D requirements compare roughly to present 
design practice in California seismic areas for buildings other than schools and hospitals. All moment 
resisting frames of concrete or steel must meet ductility requirements. Interaction effects between 
structural and nonstructural elements must be investigated. Foundation interaction requirements are 
increased. 
REFERENCES 
Bonowitz, D., N. Youssef, and J. L. Gross. 1995. A survey of steel moment-resisting frames buildings 
affected by the 1994 Northridge earthquake, Report no. NISTIR 5625, NIST, Gaithersburg, MD. 
Earthquake Engineering Research Institute. 1994. Preliminary Reconnaissance Report: Northridge 
Earthquake. 
SAC Joint Venture. 1995. Analytical Investigation of Buildings Affected by the 1994 Northridge 
Earthquake, Volumes 1 and 2, SAC 95-04A and B. SAC Joint Venture, Sacramento, CA. 
Uang, C-M. 1991. Establishing R (or Rw) and Cd Factors for Building Seismic Provisions, Journal of 
Structural Engineering, vol. 117, no. 1, pp. 19-28, ASCE. 
Wood, S. L. 1991. Performance of Reinforced Concrete Buildings During the 1985 Chile Earthquake, 
EERI Spectra, November. 

59 
In recent years, engineers and building officials have become concerned that the Provisions, and the 
building codes based on these Provisions, have become increasingly complex and difficult to understand 
and to implement. The basic driving force for this increasing complexity is the desire of the Provisions 
Update Committee to provide design guidelines that will provide for the reliable performance of 
structures. Since the response of buildings to earthquake ground shaking is by nature, very complex, 
realistic accounting for these effects leads to increasingly complex provisions. However, many of the 
current provisions have been added as prescriptive requirements relating to the design of irregularities in 
structural systems. It has been recognized that in order for buildings to be reliably constructed to resist 
earthquakes, it is necessary that the designers have sufficient understanding of the design provisions so 
that they can be properly implemented. It is feared that the typical designers of smaller, simpler 
structures, which possibly represent more than 90 percent of construction in the United States, may have 
difficulty understanding what the Provisions require in their present complex form. 
In recognition of this, as part of the BSSC 2000 Provisions Update Cycle, a special task force was 
commissioned by BSSC to develop simplified procedures, acting as an ad-hoc group reporting to TS-2. 
The approach was to develop a simplified set of the Provisions for easier application to low-rise, stiff 
structures. The procedure was designed to be used within a defined set of structures deemed to be 
sufficiently regular in configuration to allow a reduction of prescriptive requirements. The procedure 
was refined and tested over the 2000 and 2003 cycles. It is presented as a stand-alone alternate 
procedure to Chapter 4. Significant characteristics of this alternative chapter include the following: 
1. The simplified procedure would apply to structures up to three stories high in Seismic Design 
Categories B, C, D, and E, but would not be allowed for systems for which the design is typically 
controlled by considerations of drift. The task group concluded that this approach should be 
limited to certain structural systems in order to avoid problems that would arise from omitting the 
drift check for the drift-controlled systems (steel moment frames, for example). The simplified 
procedure is allowed for bearing wall and building frame systems, provided that several 
prescriptive rules are followed that result in a torsionally resistant, regular layout of lateral-loadresisting 
elements. 
2. Given the prescriptive rules for system configuration, the definitions, tables, and design 
provisions for system irregularities become unnecessary. 
3. The table of basic seismic-force-resisting systems has been shortened to include only allowable 
systems, and deflection amplification factors are not used and have been eliminated from the 
table. 
4. Design and detailing requirements have been consolidated into a single set of provisions that do 
not vary with Seismic Design Category, largely due to sections rendered unnecessary with the 
prohibition of system irregularities. 
5. The redundancy coefficient has been removed. 
6. The procedure is limited to Site Classes A to D. At the same time, it is helpful in the simplified 
method to have default Site Class Fa values for buildings and regions where detailed geotechnical 
investigations may not be available to the structural engineer. A simple definition of rock sites is 
provided in Sec. Alt. 4.6.1. As a practical matter, it should be known from a rudimentary 
geotechnical investigation whether a site is rock or soil, and so additional seismic shear wave 
velocity tests or special 100-ft. deep borings will not be necessary when utilizing this procedure. 
Alternative Simplified Chapter 4 Commentary 
2003 Commentary, Alternative Simplified Chapter 4 
60 
The default Fa values have also been set to mitigate the tendency for the SDC to be affected by 
the simplified SDS value. 
7. Vertical shear distribution is based on tributary weight. As a result, the special formula for 
calculation of diaphragm forces is removed, and calculations of diaphragm forces are greatly 
simplified. The base shear is based on the short period plateau and does not require calculation of 
the period. This base value is increased 25 percent to account for the vertical distribution method 
as well as other simplifications. A calibration study, Figure CAlt.4-1, covering a wide range of 
conditions indicates that the 25 percent adequately covers the simplifications without being overly 
conservative. 
8. Simple rigidity analysis will be required for rigid diaphragm systems, but analysis of accidental 
torsion and dynamic amplification of torsion would not be required. Untopped metal deck, wood 
panel, or plywood sheathed diaphragms may be considered flexible, representing another 
simplification in calculations. 
9. Calculations for period, drift, or P-delta effects need not be performed. 1percent drift is assumed 
when needed by requirements not covered in the simplified provisions. For example, in ACI 318, 
gravity columns are required to be designed for the calculated drift or to be specially detailed. 

Structural Design Criteria 
61 
Figure CAlt.4-1 Calibration Study. 
Calibration Study 
Simplified Lateral Force Analysis Procedure V =1.25SDS / R 
Fa = 1 for rock 1.4 for soil 
Sds = 0.67 0.93 
1.25Sds = 0.83 1.17 
Fa Values Z<=0.067 Z=0.13 Z=0.20 Z=0.27 Z>=0.33 
Site Class Ss<=0.25 Ss=0.50 Ss=0.75 Ss=1.00 Ss>=1.25 
A (hard rock) 0.80 0.80 0.80 0.80 0.80 
B (rock) 1.00 1.00 1.00 1.00 1.00 
C (soft rock) 1.20 1.20 1.10 1.00 1.00 
D (stiff soil) 1.60 1.40 1.20 1.10 1.00 
Ratio of (Simplified SDS) / (SDS = Fa x 2/3 x SS) 
Site Class Ss<=0.25 Ss=0.50 Ss=0.75 Ss=1.00 Ss>=1.25 
A (hard rock) 1.25 1.25 1.25 1.25 1.25 
B (rock) 1.00 1.00 1.00 1.00 1.00 
C (soft rock) 0.83 0.83 0.91 1.00 1.00 
D (stiff soil) 0.88 1.00 1.17 1.27 1.40 
Ratio of base shear for all buildings and overturning moment for one-story buildings 
Site Class Ss<=0.25 Ss=0.50 Ss=0.75 Ss=1.00 Ss>=1.25 
A (hard rock) 1.56 1.56 1.56 1.56 1.56 
B (rock) 1.25 1.25 1.25 1.25 1.25 
C (soft rock) 1.04 1.04 1.14 1.25 1.25 
D (stiff soil) 1.09 1.25 1.46 1.59 1.75 
Average net conservatism in overturning moment for two-story buildings 
Equal floor masses and first story height equal or up to 1.5 x second story 
Site Class Ss.<=0.25 Ss=0.50 Ss=0.75 Ss=1.00 Ss>=1.25 
A (hard rock) 1.44 1.44 1.44 1.44 1.44 
B (rock) 1.15 1.15 1.15 1.15 1.15 
C (soft rock) 0.96 0.96 1.05 1.15 1.15 
D (stiff soil) 1.01 1.15 1.34 1.46 1.61 
Average net conservatism in overturning moment for three-story 
buildings 
Equal floor masses and first story height equal or up to 1.5 x typical story 
Site Class Ss<=0.25 Ss=0.50 Ss=0.75 Ss=1.00 Ss>=1.25 
A (hard rock) 1.38 1.38 1.38 1.38 1.38 
B (rock) 1.10 1.10 1.10 1.10 1.10 
C (soft rock) 0.92 0.92 1.00 1.10 1.10 
D (stiff soil) 0.96 1.10 1.28 1.40 1.54 
Bold values indicates Seismic Design Category D in the Equivalent Lateral Force Procedure. 
Bold Italic values indicates Seismic Design Category B in the Equivalent Lateral Force Procedure. 
2003 Commentary, Alternative Simplified Chapter 4 
62 
This page is intentionally left blank. 

 63 
Chapter 5 Commentary 
STRUCTURAL ANALYSIS PROCEDURES 
5.1 GENERAL 
The equivalent lateral force (ELF) procedure specified in Sec. 5.2 is similar in its basic concept to 
SEAOC recommendations in 1968, 1973, and 1974, but several improved features have been 
incorporated. A significant revision to this procedure, which more closely reflects the ground motion 
response spectra, was adopted in the 1997 Provisions in parallel with a similar concept developed by 
SEAOC. 
The modal superposition method is a general procedure for linear analysis of the dynamic response of 
structures. In various forms, modal analysis has been widely used in the earthquake-resistant design of 
special structures such as very tall buildings, offshore drilling platforms, dams, and nuclear power plants, 
for a number of years; however, its use is also becoming more common for ordinary structures as well. 
Prior to the 1997 edition of the Provisions, the modal analysis procedure specified in Sec. 5.3 was 
simplified from the general case by restricting consideration to lateral motion in a single plane. Only one 
degree of freedom was required per floor for this type of analysis. In recent years, with the advent of 
high-speed, desktop computers, and the proliferation of relatively inexpensive, user-friendly structural 
analysis software capable of performing three dimensional modal analyses, such simplifications have 
become unnecessary. Consequently, the 1997 Provisions adopted the more general approach describing a 
three-dimensional modal analysis of the structure. When modal analysis is specified by the Provisions, a 
three-dimensional analysis generally is required except in the case of highly regular structures or 
structures with flexible diaphragms. 
The ELF procedure of Sec. 5.2 and the response spectrum procedure of Sec. 5.3 are both based on the 
approximation that the effects of yielding can be adequately accounted for by linear analysis of the 
seismic-force-resisting system for the design spectrum, which is the elastic acceleration response 
spectrum reduced by the response modification factor, R. The effects of the horizontal component of 
ground motion perpendicular to the direction under consideration in the analysis, the vertical component 
of ground motion, and torsional motions of the structure are all considered in the same simplified 
approaches in the two procedures. The main difference between the two procedures lies in the 
distribution of the seismic lateral forces over the height of the building. In the modal analysis procedure, 
the distribution is based on properties of the natural vibration modes, which are determined from the mass 
and stiffness distribution. In the ELF procedure, the distribution is based on simplified formulas that are 
appropriate for regular structures as specified in Sec. 5.2.3. Otherwise, the two procedures are subject to 
the same limitations. 
The simplifications inherent in the ELF procedure result in approximations that are likely to be 
inadequate if the lateral motions in two orthogonal directions and the torsional motion are strongly 
coupled. Such would be the case if the building were irregular in its plan configuration (see Sec. 4.3.2.2) 
or if it had a regular plan but its lower natural frequencies were nearly equal and the centers of mass and 
resistance were nearly coincident. The modal analysis method introduced in the 1997 Provisions includes 
a general model that is more appropriate for the analysis of such structures. It requires at least three 
degrees of freedom per floortwo for translational motion and one for torsional motion. 
The methods of modal analysis can be generalized further to model the effect of diaphragm flexibility, 
soil-structure interaction, etc. In the most general form, the idealization would take the form of a large 
number of mass points, each with six degrees of freedom (three translational and three rotational) 
connected by generalized stiffness elements. 
2 003 Commentary, Chapter 5 
64 
The ELF procedure (Sec. 5.2) and the response spectrum procedure are all likely to err systematically on 
the unsafe side if story strengths are distributed irregularly over height. This feature is likely to lead to 
concentration of ductility demand in a few stories of the building. The nonlinear static (or so-called 
pushover) procedure is a method to more accurately account for irregular strength distribution. However, 
it also has limitations and is not particularly applicable to tall structures or structures with relatively long 
fundamental periods of vibration. 
The actual strength properties of the various components of a structure can be explicitly considered only 
by a nonlinear analysis of dynamic response by direct integration of the coupled equations of motion. 
This method has been used extensively in earthquake research studies of inelastic structural response. If 
the two lateral motions and the torsional motion are expected to be essentially uncoupled, it would be 
sufficient to include only one degree of freedom per floor, for motion in the direction along which the 
structure is being analyzed; otherwise at least three degrees of freedom per floor, two translational and 
one torsional, should be included. It should be recognized that the results of a nonlinear response history 
analysis of such mathematical structural models are only as good as are the models chosen to represent 
the structure vibrating at amplitudes of motion large enough to cause significant yielding during strong 
ground motions. Furthermore, reliable results can be achieved only by calculating the response to several 
ground motionsrecorded accelerograms and/or simulated motionsand examining the statistics of 
response. 
It is possible with presently available computer programs to perform two- and three-dimensional inelastic 
analyses of reasonably simple structures. The intent of such analyses could be to estimate the sequence in 
which components become inelastic and to indicate those components requiring strength adjustments so 
as to remain within the required ductility limits. It should be emphasized that with the present state of the 
art in analysis, there is no one method that can be applied to all types of structures. Further, the reliability 
of the analytical results are sensitive to: 
1. The number and appropriateness of the input motion records, 
2. The practical limitations of mathematical modeling including interacting effects of inelastic elements, 
3. The nonlinear solution algorithms, and 
4. The assumed hysteretic behavior of members. 
Because of these sensitivities and limitations, the maximum base shear produced in an inelastic analysis 
should not be less than that required by Sec. 5.2. 
The least rigorous analytical procedure that may be used in determining the design seismic forces and 
deformations in structures depends on the Seismic Design Category and the structural characteristics (in 
particular, regularity). Regularity is discussed in Sec. 4.3.2. 
Except for structures assigned to Seismic Design Category A, the ELF procedure is the minimum level of 
analysis except that a more rigorous procedure is required for some Category D, E and F structures as 
identified in Table 4.4-1. The modal analysis procedure adequately addresses vertical irregularities of 
stiffness, mass, or geometry, as limited by the Provisions. Other irregularities must be carefully 
considered. 
The basis for the ELF procedure and its limitations were discussed above. It is adequate for most regular 
structures; however, the designer may wish to employ a more rigorous procedure (see list of procedures at 
beginning of this section) for those regular structures where the ELF procedure may be inadequate. The 
ELF procedure is likely to be inadequate in the following cases: 
1. Structures with irregular mass and stiffness properties in which case the simple equations for vertical 
distribution of lateral forces (Eq. 5.2-10 and 5.2-11) may lead to erroneous results; 
2. Structures (regular or irregular) in which the lateral motions in two orthogonal directions and the 
torsional motion are strongly coupled; and 
 Structural Analysis Procedures 
65 
3. Structures with irregular distribution of story strengths leading to possible concentration of ductility 
demand in a few stories of the building. 
In such cases, a more rigorous procedure that considers the dynamic behavior of the structure should be 
employed. 
Structures with certain types of vertical irregularities may be analyzed as regular structures in accordance 
with the requirements of Sec. 5.2. These structures are generally referred to as setback structures. The 
following procedure may be used: 
1. The base and tower portions of a building having a setback vertical configuration may be analyzed as 
indicated in (2) below if: 
a. The base portion and the tower portion, considered as separate structures, can be classified as 
regular and 
b. The stiffness of the top story of the base is at least five times that of the first story of the tower. 
When these conditions are not met, the building must be analyzed in accordance with Sec. 5.3. 
2. The base and tower portions may be analyzed as separate structures in accordance with the following: 
a. The tower may be analyzed in accordance with the procedures in Sec. 5.2 with the base taken at 
the top of the base portion. 
b. The base portion then must be analyzed in accordance with the procedures in Sec. 5.2 using the 
height of the base portion of hn and with the gravity load and seismic base shear forces of the 
tower portion acting at the top level of the base portion. 
The design requirements in Sec. 5.3 include a simplified version of modal analysis that accounts for 
irregularity in mass and stiffness distribution over the height of the building. It would be adequate, in 
general, to use the ELF procedure for structures whose floor masses and cross-sectional areas and 
moments of inertia of structural members do not differ by more than 30 percent in adjacent floors and in 
adjacent stories. 
For other structures, the following procedure should be used to determine whether the modal analysis 
procedures of Sec. 5.3 should be used: 
1. Compute the story shears using the ELF procedure specified in Sec. 5.2. 
2. On this basis, approximately dimension the structural members, and then compute the lateral 
displacements of the floor. 
3. Replace h in Eq. 5.2-11 with these displacements, and recompute the lateral forces to obtain the 
revised story shears. 
4. If at any story the recomputed story shear differs from the corresponding value as obtained from the 
procedures of Sec. 5.2 by more than 30 percent, the building should be analyzed using the procedure 
of Sec. 5.3. If the difference is less than this value, the building may be designed using the story 
shear obtained in the application of the present criterion and the procedures of Sec. 5.3 are not 
required. 
Application of this procedure to these structures requires far less computational effort than the use of the 
response spectrum procedure of Sec. 5.3. In the majority of the structures, use of this procedure will 
determine that modal analysis need not be used and will also furnish a set of story shears that practically 
always lie much closer to the results of modal analysis than the results of the ELF procedure. 
This procedure is equivalent to a single cycle of Newmarks method for calculation of the fundamental 
mode of vibration. It will detect both unusual shapes of the fundamental mode and excessively high 
influence of higher modes. Numerical studies have demonstrated that this procedure for determining 
whether modal analysis must be used will, in general, detect cases that truly should be analyzed 
2 003 Commentary, Chapter 5 
66 
dynamically; however, it generally will not indicate the need for dynamic analysis when such an analysis 
would not greatly improve accuracy. 
5.2 EQUIVALENT LATERAL FORCE PROCEDURE 
This section discusses the equivalent lateral force (ELF) procedure for seismic analysis of structures. 
5.2.1 Seismic base shear. The heart of the ELF procedure is Eq. 5.2-1 for base shear, which gives the 
total seismic design force, V, in terms of two factors: a seismic response coefficient, Cs, and the seismic 
weight, W. The seismic response coefficient Cs, is obtained from Eq. 5.2-2 and 5.2-3 based on the design 
spectral response acceleration parameters, SDS and SD1. These acceleration parameters and the derivation 
of the response spectrum is discussed more fully in the Commentary for Chapter 3. The seismic weight is 
discussed in Commentary Sec. 1.5.1. 
The base shear formula and the various factors contained therein were arrived at as explained below. 
Elastic acceleration response spectrum. See the Commentary to Chapter 4 for a full discussion of the 
shape of the spectrum accounting for dynamic response amplification and the effect of site response. 
Elastic design spectrum. The elastic acceleration response spectrum for earthquake motions has a 
descending branch for longer values of T, the period of vibration of the system, that varies roughly as a 
function of 1/T. In previous editions of the Provisions, the actual response spectra that varied in a 1/T 
relationship were replaced with design spectra that varied in a 1/T2/3 relationship. This was intentionally 
done to provide added conservatism in the design of tall structures, as well as to account for the effects of 
higher mode participation. In the development of the 1997 Provisions, a special task force, known as the 
Seismic Design Procedures Group (SDPG), was convened to develop a method for using new seismic 
hazard maps, developed by the USGS in the Provisions. Whereas older seismic hazard maps provided an 
effective peak ground acceleration coefficient, Ca, and an effective peak velocity-related acceleration 
coefficient, Cv, the new maps directly provide parameters that correspond to points on the response 
spectrum. It was the recommendation of the SDPG that the true shape of the response spectrum, 
represented by a 1/T relationship, be used in the base shear equation. In order to maintain the added 
conservatism for tall and high occupancy structures, formerly provided by the design spectra which 
utilized a 1/T2/3 relationship, the 1997 Provisions adopted an occupancy importance factor I into the base 
shear equation. This I factor, which has a value of 1.25 for Seismic Use Group II structures and 1.5 for 
Seismic Use Group III structures has the effect of raising the design spectrum for taller, high occupancy 
structures, to levels comparable to those for which they were designed in previous editions of the 
Provisions. 
Although the introduction of an occupancy importance factor in the 1997 edition adjusted the base shear 
to more conservative values for large buildings with higher occupancies, it did not address the issue of 
accounting for higher mode effects, which can be significant in longer period structuresthose with 
fundamental modes of vibration significantly larger than the period Ts, at which the response spectrum 
changes from one of constant response acceleration (Eq. 5.2-2) to one of constant response velocity 
(Eq. 5.2-3). 
Equation 5.2-3 could be modified to produce an estimate of base shear that is more consistent with the 
results predicted by elastic response spectrum methods. Some suggestions for such modifications may be 
found in Chopra (1995). However, it is important to note that even if the base shear equation were to 
simulate results of an elastic response spectrum analysis more accurately, most structures respond to 
design level ground shaking in an inelastic manner. This inelastic response results in different demands 
than are predicted by elastic analysis, regardless of how exact the analysis is. Inelastic response 
behavior in multistory buildings could be partially accounted for by other modifications to the seismic 
coefficient Cs. Specifically, the coefficient could be made larger to limit the ductility demand in 
multistory buildings to the same value as for single-degree-of-freedom systems. Results supporting such 
an approach may be found in (Chopra, 1995) and in (Nassar and Krawinkler, 1991). 
 Structural Analysis Procedures 
67 
The above notwithstanding, the equivalent lateral force procedure is intended to provide a relatively 
straightforward design approach where complex analyses, accurately accounting for dynamic and 
inelastic response effects, are not warranted. Rather than making the procedure more complex, so that it 
would be more appropriate for structures with significant higher mode response, in the 2000 edition of the 
Provisions application of this technique to structures assigned to Seismic Design Categories D, E, and F is 
limited to those where higher mode effects are not significant. Given the widespread use of computerassisted 
analysis for major structures, it was felt that these limitations on the application of the equivalent 
lateral force procedure would not be burdensome. It should be noted that particularly for tall structures, 
the use of dynamic analysis methods will not only result in a more realistic characterization of the 
distribution of inertial forces in the structure, but may also result in reduced forces, particularly with 
regard to overturning demands. Therefore, use of a dynamic analysis method is recommended for such 
structures, regardless of the Seismic Design Category. 
Historically, the ELF analytical approach has been limited in application in Seismic Design Categories D, 
E, and F to regular structures with heights of 240 ft (70 m) or less and irregular structures with heights of 
100 ft (30 m) or less. Following recognition that the use of a base shear equation with a 1/T relationship 
underestimated the response of structures with significant higher mode participation, a change in the 
height limit for regular structures to 100 ft (30 m) was contemplated. However, the importance of higher 
mode participation in structural response is a function both of the structures dynamic properties, which 
are dependent on height, mass and the stiffness of various lateral force resisting elements, and of the 
frequency content of the ground shaking, as represented by the response spectrum. Therefore, rather than 
continuing to use building height as the primary parameter used to control analysis procedures, it was 
decided to limit the application of the ELF to those structures in Seismic Design Categories D, E, and F 
having fundamental periods of response less than 3.5 times the period at which the response spectrum 
transitions from constant response acceleration to constant response velocity. This limit was selected 
based on comparisons of the base shear calculated by the ELF equations to that predicted by response 
spectrum analysis for structures of various periods on five different sites, representative of typical 
conditions in the eastern and western United States. For all 5 sites, it was determined that the ELF 
equations conservatively bound the results of a response spectrum analysis for structures having periods 
lower than the indicated amount. 
Figure C5.2-2 Measured building period for momentresisting 
steel frame structures. 
2 003 Commentary, Chapter 5 
68 
Response modification factor. The factor R in the denominator of Eq. 5.2-2 and 5.2-3 is an empirical 
response reduction factor intended to account for damping, overstrength, and the ductility inherent in the 
structural system at displacements great enough to surpass initial yield and approach the ultimate load 
displacement of the structural system. Thus, for a lightly damped building structure of brittle material 
that would be unable to tolerate any appreciable deformation beyond the elastic range, the factor R would 
be close to 1 (that is, no reduction from the linear elastic response would be allowed). At the other 
extreme, a heavily damped building structure with a very ductile structural system would be able to 
withstand deformations considerably in excess of initial yield and would, therefore, justify the assignment 
of a larger response reduction factor R. Table 4.3-1 in the Provisions stipulates R factors for different 
types of building systems using several different structural materials. The coefficient R ranges in value 
from a minimum of 1 for an unreinforced masonry bearing wall system to a maximum of 8 for a special 
moment frame system. The basis for the R factor values specified in Table 4.3-1 is explained in the 
Commentary to Sec. 4.2.1. 
The effective value of R used in the base shear equation is adjusted by the occupancy importance factor I. 
The value of I, which ranges from 1 to 1.5, has the effect of reducing the amount of ductility the structure 
will be called on to provide at a given level of ground shaking. However, it must be recognized that 
added strength, by itself, is not adequate to provide for superior seismic performance in buildings with 
critical occupancies. Good connections and construction details, quality assurance procedures, and 
limitations on building deformation or drift are also important to significantly improve the capability for 
maintenance of function and safety in critical facilities and those with a high-density occupancy. 
Consequently, the reduction in the damage potential of critical facilities (Group III) is also handled by 
using more conservative drift controls (Sec. 4.5.1) and by providing special design and detailing 
requirements (Sec. 4.6) and materials limitations (Chapters 8 through 12). 
5.2.2 Period determination. In the denominator of Eq. 5.2-3, T is the fundamental period of vibration of 
the structure. It is preferable that this be determined using modal analysis methods and the principles of 
structural mechanics. However, methods of structural mechanics cannot be employed to calculate the 
vibration period before a structure has been designed. Consequently, this section provides an 
approximate method that can be used to estimate the period, with minimal information available on the 
design. It is based on the use of simple formulas that involve only a general description of the type of 
Figure C5.2-1 Measured building period for 
reinforced concrete frame structures.
 Structural Analysis Procedures 
69 
structure (such as steel moment frame, concrete moment frame, shear wall system, braced frame) and 
overall dimensions (such as height and plan length) to estimate the period of vibration in order to 
calculate an initial base shear and proceed with a preliminary design. 
It is advisable that this base shear and the corresponding value of T be conservative. 
Even for final design, use of an unrealistically large value for T is unconservative. Thus, the value of T 
used in design should be smaller than the period calculated for the bare frame of the building. Equations 
5.2-6, 5.2-7, and 5.2-8 for the approximate period Ta are therefore intended to provide conservative 
estimates of the fundamental period of vibration. An upper bound is placed on the value of T calculated 
using more exact methods, based on Ta and the factor Cu. The coefficient Cu is intended to reflect the 
likelihood that buildings in areas with lower lateral force requirements probably will be more flexible. 
Furthermore, it results in less dramatic changes from present practice in lower risk areas. It is generally 
accepted that the empirical equations for Ta are tailored to fit the type of construction common in areas 
with high lateral force requirements. It is unlikely that buildings in lower risk seismic areas would be 
designed to produce as high a drift level as allowed in the Provisions due to stability problems (P-delta) 
and wind requirements. Where the design of a structure is actually controlled by wind, the use of a 
large T will not really result in a lower design force; thus, use of this approach in high-wind regions 
should not result in unsafe design. 
Taking the seismic base shear to vary as a function of 1/T and assuming that the lateral forces are 
distributed linearly over the height and that the deflections are controlled by drift limitations, a simple 
calculation of the period of vibration by Rayleighs method leads to the conclusion that the vibration 
period of moment resisting frame structures varies roughly as hn
3/4 where hn equals the total height of the 
building as defined elsewhere. Based on this, for many years Eq. 5.2-6 appeared in the Provisions in the 
form: 
3/4 
Ta=Cthn 
Figure C5.2-3 Measured building period for concrete shear 
wall structures. 
2 003 Commentary, Chapter 5 
70 
A large number of strong motion instruments have been placed in buildings located within zones of high 
seismic activity by the U.S. Geological Survey and the California Division of Mines and Geology. Over 
the past several years, this has allowed the response to strong ground shaking for a significant number of 
these buildings to be recorded and the fundamental period of vibration of the buildings to be calculated. 
Figures C5.2-1, C5.2-2, and C5.2-3, respectively, show plots of these data as a function of building height 
for three classes of structures. Figure C5.2-1 shows the data for moment-resisting concrete frame 
buildings; Figure C5.2-2, for moment-resisting steel frame buildings; and Figure C5.2-3, for concrete 
shear wall buildings. Also shown in these figures are equations for lines that envelop the data within 
approximately a standard deviation above and below the mean. 
For the 2000 Provisions, Eq. 5.2-6 is revised into a more general form allowing the statistical fits of the 
data shown in the figures to be used directly. The values of the coefficient Cr and the exponent x given in 
Table 5.2-2 for these moment-resisting frame structures represent the lower bound (mean minus one 
standard deviation) fits to the data shown in Figures C5.2-1 and C5.2-2, respectively, for steel and 
concrete moment frames. Although updated data were available for concrete shear wall structures, these 
data do not fit well with an equation of the form of Eq. 5.2-6. This is because the period of shear wall 
buildings is highly dependent not only on the height of the structure but also on the amount of shear wall 
present in the building. Analytical evaluations performed by Chopra and Goel (1997 and 1998) indicate 
that equations of the form of Eq. 5.2-8 and 5.2-9 provide a reasonably good fit to the data. However, the 
form of these equations is somewhat complex. Therefore, the simpler form of Eq. 5.2-6 contained in 
earlier editions of the Provisions was retained with the newer, more accurate formulation presented as an 
alternative. 
Updated data for other classes of construction were not available. As a result, the Cr and x values for 
other types of construction shown in Table 5.2-2 are values largely based on limited data obtained from 
the 1971 San Fernando earthquake that have been used in past editions of the Provisions. The optional 
use of T = 0.1N (Eq. 5.2-7) is an approximation for low to moderate height frames that has long been in 
use. 
In earlier editions of the Provisions, the Cu coefficient varied from a value of 1.2 in zones of high 
seismicity to a value of 1.7 in zones of low seismicity. The data presented in Figures C5.2-1, C5.2-2, and 
C5.2-3 permit direct evaluation of the upper bound on period as a function of the lower bound, given by 
Eq. 5.2-6. This data indicates that in zones of high seismicity, the ratio of the upper to lower bound may 
more properly be taken as a value of about 1.4. Therefore, in the 2000 Provisions, the values in 
Table 5.2-1 were revised to reflect this data in zones of high seismicity while retaining the somewhat 
subjective values contained in earlier editions for the zones of lower seismicity. 
For exceptionally stiff or light buildings, the calculated T for the seismic-force-resisting system may be 
significantly shorter than Ta calculated by Eq. 5.2-6. For such buildings, it is recommended that the 
period value T be used in lieu of Ta for calculating the seismic response coefficient, Cs. 
Although the approximate methods of Sec. 5.2.2.1 can be used to determine a period for the design of 
structures, the fundamental period of vibration of the seismic-force-resisting system should be calculated 
according to established methods of mechanics. Computer programs are available for such calculations. 
One method of calculating the period, probably as convenient as any, is the use of the following formula 
based on Rayleighs method: 
2 
1 
1 
2 
n 
i i 
i 
n 
i i 
i 
w 
T 
g F
d 
p 
d 
= 
= 
S 
= 
S 
(C5.2-1) 
where: 
Fi = the seismic lateral force at Level i, 
 Structural Analysis Procedures 
71 
wi = the seismic weight assigned in Level i, 
di 
= the static lateral displacement at Level i due to the forces Fi computed on a linear elastic 
basis, and 
g = is the acceleration due to gravity. 
The calculated period increases with an increase in flexibility of the structure because the d term in the 
Rayleigh formula appears to the second power in the numerator but to only the first power in the 
denominator. Thus, if one ignores the contribution of nonstructural elements to the stiffness of the 
structure in calculating the deflections d, the deflections are exaggerated and the calculated period is 
lengthened, leading to a decrease in the seismic response coefficient Cs and, therefore, a decrease in the 
design force. Nonstructural elements participate in the behavior of the structure even though the designer 
may not rely on them to contribute any strength or stiffness to the structure. To ignore them in 
calculating the period is to err on the unconservative side. The limitation of CuTa is imposed as a 
safeguard. 
5.2.3 Vertical distribution of seismic forces. The distribution of lateral forces over the height of a 
structure is generally quite complex because these forces are the result of superposition of a number of 
natural modes of vibration. The relative contributions of these vibration modes to the total forces depends 
on a number of factors including the shape of the earthquake response spectrum, the natural periods of 
vibration of the structure, and the shapes of vibration modes that, in turn, depend on the distribution of 
mass and stiffness over the height. The basis of this method is discussed below. In structures having only 
minor irregularity of mass or stiffness over the height, the accuracy of the lateral force distribution as 
given by Eq. 5.2-11 is much improved by the procedure described in the last portion of Sec. 5.1 of this 
commentary. The lateral force at each level, x, due to response in the first (fundamental) natural mode of 
vibration is given by Eq. C5.2-2 as follows: 
1 
1 1 
1 1 
1
x x 
x n 
i 
i 
f V w
w
f
f 
= 
. . 
. . 
= . . 
...S ... 
(C5.2-2) 
where: 
V1 = the contribution of this mode to the base shear, 
wi = the weight lumped at the ith level, and 
fi 
= the amplitude of the first mode at the ith level. 
This is the same as Eq. 5.3-7 in Sec. 5.3.5 of the Provisions, but it is specialized for the first mode. If V1 
is replaced by the total base shear, V, this equation becomes identical to Eq. 5.2-11 with k = 1 if the first 
mode shape is a straight line and with k = 2 if the first mode shape is a parabola with its vertex at the base. 
It is well known that the influence of modes of vibration higher than the fundamental mode is small in the 
earthquake response of short period structures and that, in regular structures, the fundamental vibration 
mode departs little from a straight line. This, along with the matters discussed above, provides the basis 
for Eq. 5.2-11 with k = 1 for structures having a fundamental vibration period of 0.5 seconds or less. 
It has been demonstrated that although the earthquake response of long period structures is primarily due 
to the fundamental natural mode of vibration, the influence of higher modes of vibration can be 
significant and, in regular structures, the fundamental vibration mode lies approximately between a 
straight line and a parabola with the vertex at the base. Thus, Eq. 5.2-11 with k = 2 is appropriate for 
structures having a fundamental period of vibration of 2.5 seconds or longer. Linear variation of k 
between 1 at a 0.5 second period and 2 at a 2.5 seconds period provides the simplest possible transition 
between the two extreme values. 
2 003 Commentary, Chapter 5 
72 
5.2.4 Horizontal shear distribution. The story shear in any story is the sum of the lateral forces acting 
at all levels above that story. Story x is the story immediately below Level x (Figure C5.2-4). Reasonable 
and consistent assumptions regarding the stiffness of concrete and masonry elements may be used for 
analysis in distributing the shear force to such elements connected by a horizontal diaphragm. Similarly, 
the stiffness of moment or braced frames will establish the distribution of the story shear to the vertical 
resisting elements in that story. 
5.2.4.1 and 5.2.4.2 Inherent and accidental torsion. The torsional moment to be considered in the 
design of elements in a story consists of two parts: 
1. Mt, the moment due to eccentricity between centers of mass and resistance for that story, which is 
computed as the story shear times the eccentricity perpendicular to the direction of applied earthquake 
forces. 
2. Mta, commonly referred to as accidental torsion, which is computed as the story shear times the 
accidental eccentricity, equal to 5 percent of the dimension of the structure (in the story under 
consideration) perpendicular to the direction of the applied earthquake forces. 
Computation of Mta in this manner is equivalent to the procedure in Sec. 5.2.4.2 which implies that the 
dimension of the structure is the dimension in the story where the torsional moment is being computed 
and that all the masses above that story should be assumed to be displaced in the same direction at one 
time (for example, first, all of them to the left and, then, to the right). 
Dynamic analyses assuming linear behavior indicate that the torsional moment due to eccentricity 
between centers of mass and resistance may significantly exceed Mt (Newmark and Rosenblueth, 1971). 
However, such dynamic magnification is not included in the Provisions, partly because its significance is 
not well understood for structures designed to deform well beyond the range of linear behavior. 
Figure C5.2-4 Description of story and level. 
The shear at Story x (Vx) is the sum of all the lateral forces at 
and above Story x (Fx through Fn). 
 Structural Analysis Procedures 
73 
The torsional moment Mt calculated in accordance with this provision would be zero in those stories 
where centers of mass and resistance coincide. However, during vibration of the structure, torsional 
moments would be induced in such stories due to eccentricities between centers of mass and resistance in 
other stories. To account for such effects, it is recommended that the torsional moment in any story be no 
smaller than the following two values (Newmark and Rosenblueth, 1971): 
1. The story shear times one-half of the maximum of the computed eccentricities in all stories below the 
one being analyzed and 
2. One-half of the maximum of the computed torsional moments for all stories above. 
Accidental torsion is intended to cover the effects of several factors that have not been explicitly 
considered in the Provisions. These factors include the rotational component of ground motion about a 
vertical axis; unforeseeable differences between computed and actual values of stiffness, yield strengths, 
and dead-load masses; and unforeseeable unfavorable distributions of dead- and live-load masses. 
The way in which the story shears and the effects of torsional moments are distributed to the vertical 
elements of the seismic-force-resisting system depends on the stiffness of the diaphragms relative to 
vertical elements of the system. 
Where the diaphragm stiffness in its own plane is sufficiently high relative to the stiffness of the vertical 
components of the system, the diaphragm may be assumed to be indefinitely rigid for purposes of this 
section. Then, in accordance with compatibility and equilibrium requirements, the shear in any story is to 
be distributed among the vertical components in proportion to their contributions to the lateral stiffness of 
the story while the story torsional moment produces additional shears in these components that are 
proportional to their contributions to the torsional stiffness of the story about its center of resistance. This 
contribution of any component is the product of its lateral stiffness and the square of its distance to the 
center of resistance of the story. Alternatively, the story shears and torsional moments may be distributed 
on the basis of a three-dimensional analysis of the structure, consistent with the assumption of linear 
behavior. 
Where the diaphragm in its own plane is very flexible relative to the vertical components, each vertical 
component acts nearly independently of the rest. The story shear should be distributed to the vertical 
components considering these to be rigid supports. Analysis of the diaphragm acting as a continuous 
horizontal beam or truss on rigid supports leads to the distribution of shears. Because the properties of 
the beam or truss may not be accurately computed, the shears in vertical elements should not be taken to 
be less than those based on tributary areas. Accidental torsion may be accounted for by adjusting the 
position of the horizontal force with respect to the supporting vertical elements. 
There are some common situations where it is obvious that the diaphragm can be assumed to be either 
rigid or very flexible in its own plane for purposes of distributing story shear and considering torsional 
moments. For example, a solid monolithic reinforced concrete slab, square or nearly square in plan, in a 
structure with slender moment resisting frames may be regarded as rigid. A large plywood diaphragm 
with widely spaced and long, low masonry walls may be regarded as very flexible. In intermediate 
situations, the design forces should be based on an analysis that explicitly considers diaphragm 
deformations and satisfies equilibrium and compatibility requirements. Alternatively, the design forces 
could be based on the envelope of the two sets of forces resulting from both extreme assumptions 
regarding the diaphragmsrigid or very flexible. 
Where the horizontal diaphragm is not continuous and the elements perpendicular to the direction of 
motion are ignored, the story shear can be distributed to the vertical components based on their tributary 
areas. 
5.2.4.3 Dynamic amplification of torsion. There are indications that the 5 percent accidental 
eccentricity may be too small in some structures since they may develop torsional dynamic instability. 
Some examples are the upper stories of tall structures having little or no nominal eccentricity, those 
structures where the calculations of relative stiffnesses of various elements are particularly uncertain 
(such as those that depend largely on masonry walls for lateral force resistance or those that depend on 
2 003 Commentary, Chapter 5 
74 
vertical elements made of different materials), and nominally symmetrical structures that utilize core 
elements alone for seismic resistance or that behave essentially like elastic nonlinear systems (for 
example, some prestressed concrete frames). The amplification factor for torsionally irregular structures 
(Eq. 5.2-13) was introduced in the 1988 Edition as an attempt to account for some of these problems in a 
controlled and rational way. 
5.2.5 Overturning. This section requires that the structure be designed to resist overturning moments 
statically consistent with the design story shears. In the 1997 and earlier editions of the Provisions, the 
overturning moment was modified by a factor, t, to account, in an approximate manner, for the effects of 
higher mode response in taller structures. In the 2000 edition of the Provisions, the equivalent lateral 
force procedure was limited in application in Seismic Design Categories D, E, and F to structures that do 
not have significant higher mode participation. As a result it was possible to simplify the design 
procedure by eliminating the t factor. Under this new approach tall structures in Seismic Design 
Categories B and C designed using the equivalent lateral force procedure will be designed for somewhat 
larger overturning demands than under past editions of the Provisions. This conservatism was accepted 
as an inducement for designers of such structures to use a more appropriate dynamic analysis procedure. 
In the design of the foundation, the overturning moment calculated at the foundation-soil interface may be 
reduced to 75 percent of the calculated value using Eq. 5.2-14. This is appropriate because a slight 
uplifting of one edge of the foundation during vibration leads to reduction in the overturning moment and 
because such behavior does not normally cause structural distress. 
5.2.6 Drift determination and P-delta effects. This section defines the design story drift as the 
difference of the deflections, dx, at the top and bottom of the story under consideration. The deflections, 
dx, are determined by multiplying the deflections, dxe (determined from an elastic analysis), by the 
deflection amplification factor, Cd, given in Table 4.3-1. The elastic analysis is to be made for the 
seismic-force-resisting system using the prescribed seismic design forces and considering the structure to 
be fixed at the base. Stiffnesses other than those of the seismic-force-resisting system should not be 
included since they may not be reliable at higher inelastic strain levels. 
The deflections are to be determined by combining the effects of joint rotation of members, shear 
deformations between floors, the axial deformations of the overall lateral resisting elements, and the shear 
and flexural deformations of shear walls and braced frames. The deflections are determined initially on 
the basis of the distribution of lateral forces stipulated in Sec. 5.2.3. For frame structures, the axial 
deformations from bending effects, although contributing to the overall structural distortion, may or may 
not affect the story-to-story drift; however, they are to be considered. Centerline dimensions between the 
frame elements often are used for analysis, but clear-span dimensions with consideration of joint panel 
zone deformation also may be used. 
For determining compliance with the story drift limitation of Sec. 4.5.1, the deflections, dx, may be 
calculated as indicated above for the seismic-force-resisting system and design forces corresponding to 
the fundamental period of the structure, T (calculated without the limit T = CuTa specified in Sec. 5.2.2), 
may be used. The same model of the seismic-force-resisting system used in determining the deflections 
must be used for determining T. The waiver does not pertain to the calculation of drifts for determining 
P-delta effects on member forces, overturning moments, etc. If the P-delta effects determined in Sec. 
5.2.6.2 are significant, the design story drift must be increased by the resulting incremental factor. 
The P-delta effects in a given story are due to the eccentricity of the gravity load above that story. If the 
story drift due to the lateral forces prescribed in Sec. 5.2.3 were ., the bending moments in the story 
would be augmented by an amount equal to . times the gravity load above the story. The ratio of the 
P-delta moment to the lateral force story moment is designated as a stability coefficient, ., in Eq. 5.2-16. 
If the stability coefficient . is less than 0.10 for every story, the P-delta effects on story shears and 
moments and member forces may be ignored. If, however, the stability coefficient . exceeds 0.10 for any 
story, the P-delta effects on story drifts, shears, member forces, etc., for the whole structure must be 
determined by a rational analysis. 
An acceptable P-delta analysis, based upon elastic stability theory, is as follows: 
 Structural Analysis Procedures 
75 
1. Compute for each story the P-delta amplification factor, ad = ./(1 - .). ad takes into account the 
multiplier effect due to the initial story drift leading to another increment of drift that would lead to 
yet another increment, etc. Thus, both the effective shear in the story and the computed eccentricity 
would be augmented by a factor 1 + . + . 2 + . 3 ..., which is 1/(1 - .) or (1 + ad). 
2. Multiply the story shear, Vx, in each story by the factor (1 + ad) for that story and recompute the story 
shears, overturning moments, and other seismic force effects corresponding to these augmented story 
shears. 
This procedure is applicable to planar structures and, with some extension, to three-dimensional 
structures. Methods exist for incorporating two- and three-dimensional P-delta effects into computer 
analyses that do not explicitly include such effects (Rutenburg, 1985). Many programs explicitly include 
P-delta effects. A mathematical description of the method employed by several popular programs is 
given by Wilson and Habibullah (1987). 
The P-delta procedure cited above effectively checks the static stability of a structure based on its initial 
stiffness. Since the inception of this procedure with ATC 3-06, however, there has been some debate 
regarding its accuracy. This debate stems from the intuitive notion that the structures secant stiffness 
would more accurately represent inelastic P-delta effects. Given the additional uncertainty of the effect of 
dynamic response on P-delta behavior and the (apparent) observation that instability-related failures 
rarely occur in real structures, the P-delta requirements remained as originally written until revised for the 
1991 Edition. 
There was increasing evidence that the use of elastic stiffness in determining theoretical P-delta response 
is unconservative. Given a study carried out by Bernal (1987), it was argued that P-delta amplifiers 
should be based on secant stiffness and that, in other words, the Cd term in Eq. 5.2-16 should be deleted. 
However, since Bernals study was based on the inelastic response of single-degree-of-freedom, elasticperfectly 
plastic systems, significant uncertainties existed regarding the extrapolation of the concepts to 
the complex hysteretic behavior of multi-degree-of-freedom systems. 
Another problem with accepting a P-delta procedure based on secant stiffness is that design forces would 
be greatly increased. For example, consider an ordinary moment frame of steel with a Cd of 4.0 and an 
elastic stability coefficient . of 0.15. The amplifier for this structure would be 1.0/0.85 = 1.18 according 
to the 1988 Edition of the Provisions. If the P-delta effects were based on secant stiffness, however, the 
stability coefficient would increase to 0.60 and the amplifier would become 1.0/0.4 = 2.50. This example 
illustrates that there could be an extreme impact on the requirements if a change were implemented that 
incorporated P-delta amplifiers based on static secant stiffness response. 
There was, however, some justification for retaining the P-delta amplifier as based on elastic stiffness. 
This justification was the apparent lack of stability-related failures. The reasons for the lack of observed 
failures included: 
1. Many structures display strength well above the strength implied by code-level design forces (see 
Figure C4.2-3). This overstrength likely protects structures from stability-related failures. 
2. The likelihood of a failure due to instability decreases with increased intensity of expected groundshaking. 
This is due to the fact that the stiffness of most structures designed for extreme ground 
motion is significantly greater than the stiffness of the same structure designed for lower intensity 
shaking or for wind. Since damaging, low-intensity earthquakes are somewhat rare, there would be 
little observable damage. 
Due to the lack of stability-related failures, therefore, recent editions of the Provisions regarding P-delta 
amplifiers have remained from the 1991 Editions. 
The 1991 Edition introduced a requirement that the computed stability coefficient, ., not exceed 0.25 or 
0.5/Cd, where Cd is an adjusted ductility demand that takes into account the fact that the seismic 
strength demand may be somewhat less than the code strength supplied. The adjusted ductility demand is 
2 003 Commentary, Chapter 5 
76 
not intended to incorporate overstrength beyond that computed by the means available in Chapters 8 
through 14 of the Provisions. 
The purpose of this requirement is to protect structures from the possibility of stability failures triggered 
by post-earthquake residual deformation. The danger of such failures is real and may not be eliminated 
by apparently available overstrength. This is particularly true of structures designed in regions of lower 
seismicity. 
The computation of .max, which, in turn, is based on Cd, requires the computation of story strength 
supply and story strength demand. Story strength demand is simply the seismic design shear for the story 
under consideration. The story strength supply may be computed as the shear in the story that occurs 
simultaneously with the attainment of the development of first significant yield of the overall structure. 
To compute first significant yield, the structure should be loaded with a seismic force pattern similar to 
that used to compute seismic story strength demand. A simple and conservative procedure is to compute 
the ratio of demand to strength for each member of the seismic-force-resisting system in a particular story 
and then use the largest such ratio as . For a structure otherwise in conformance with the Provisions, 
taking  equal to 1.0 is obviously conservative. 
The principal reason for inclusion of  is to allow for a more equitable analysis of those structures in 
which substantial extra strength is provided, whether as a result of added stiffness for drift control, for 
code-required wind resistance, or simply a feature of other aspects of the design. Some structures 
inherently possess more strength than required, but instability is not typically a concern for such 
structures. For many flexible structures, the proportions of the structural members are controlled by the 
drift requirements rather than the strength requirements; consequently,  is less than 1.0 because the 
members provided are larger and stronger than required. This has the effect of reducing the inelastic 
component of total seismic drift and, thus,  is placed as a factor on Cd. 
Accurate evaluation of  would require consideration of all pertinent load combinations to find the 
maximum value of seismic load effect demand to seismic load effect capacity in each and every member. 
A conservative simplification is to divide the total demand with seismic included by the total capacity; 
this covers all load combinations in which dead and live effects add to seismic. If a member is controlled 
by a load combination where dead load counteracts seismic, to be correctly computed, the ratio  must be 
based only on the seismic component, not the total; note that the vertical load P in the P-delta 
computation would be less in such a circumstance and, therefore, . would be less. The importance of the 
counteracting load combination does have to be considered, but it rarely controls instability. 
5.3 RESPONSE SPECTRUM PROCEDURE 
Modal analysis (Newmark and Rosenblueth, 1971; Clough and Penzien, 1975; Thomson, 1965; Wiegel, 
1970) is applicable for calculating the linear response of complex, multi-degree-of-freedom structures and 
is based on the fact that the response is the superposition of the responses of individual natural modes of 
vibration, each mode responding with its own particular pattern of deformation (the mode shape), with its 
own frequency (the modal frequency), and with its own modal damping. The response of the structure, 
therefore, can be modeled by the response of a number of single-degree-of-freedom oscillators with 
properties chosen to be representative of the mode and the degree to which the mode is excited by the 
earthquake motion. For certain types of damping, this representation is mathematically exact and, for 
structures, numerous full-scale tests and analyses of earthquake response of structures have shown that 
the use of modal analysis, with viscously damped single-degree-of-freedom oscillators describing the 
response of the structural modes, is an accurate approximation for analysis of linear response. 
Modal analysis is useful in design. The ELF procedure of Sec. 5.2 is simply a first mode application of 
this technique, which assumes all of the structures mass is active in the first mode. The purpose of 
modal analysis is to obtain the maximum response of the structure in each of its important modes, which 
are then summed in an appropriate manner. This maximum modal response can be expressed in several 
ways. For the Provisions, it was decided that the modal forces and their distributions over the structure 
should be given primary emphasis to highlight the similarity to the equivalent static methods traditionally 
 Structural Analysis Procedures 
77 
used in building codes (the SEAOC recommendations and the UBC) and the ELF procedure in Sec. 5.2. 
Thus, the coefficient Csm in Eq. 5.3-3 and the distribution equations, Eq. 5.3-1 and 5.3-2, are the 
counterparts of Eq. 5.2-10 and 5.2-11. This correspondence helps clarify the fact that the simplified 
modal analysis contained in Sec. 5.3 is simply an attempt to specify the equivalent lateral forces on a 
structure in a way that directly reflects the individual dynamic characteristics of the structure. Once the 
story shears and other response variables for each of the important modes are determined and combined to 
produce design values, the design values are used in basically the same manner as the equivalent lateral 
forces given in Sec. 5.2. 
5.3.2 Modes. This section defines the number of modes to be used in the analysis. For many structures, 
including low-rise structures and structures of moderate height, three modes of vibration in each direction 
are nearly always sufficient to determine design values of the earthquake response of the structure. For 
high-rise structures, however, more than three modes may be required to adequately determine the forces 
for design. This section provides a simple rule that the combined participating mass of all modes 
considered in the analysis should be equal to or greater than 90 percent of the effective total mass in each 
of two orthogonal horizontal directions. 
5.3.3 Modal properties. Natural periods of vibration are required for each of the modes used in the 
subsequent calculations. These are needed to determine the modal coefficients Csm is Sec. 5.3.4. Because 
the periods of the modes contemplated in these requirements are those associated with moderately large, 
but still essentially linear, structural response, the period calculations should include only those elements 
that are effective at these amplitudes. Such periods may be longer than those obtained from a 
small-amplitude test of the structure when completed or the response to small earthquake motions because 
of the stiffening effects of nonstructural and architectural components of the structure at small amplitudes. 
During response to strong ground-shaking, however, measured responses of structures have shown that 
the periods lengthen, indicating the loss of the stiffness contributed by those components. 
There exists a wide variety of methods for calculation of natural periods and associated mode shapes, and 
no one particular method is required by the Provisions. It is essential, however, that the method used be 
one based on generally accepted principles of mechanics such as those given in well known textbooks on 
structural dynamics and vibrations (Clough and Penzien, 1975; Newmark and Rosenblueth, 1971; 
Thomson, 1965; Wiegel, 1970). Although it is expected that in many cases computer programs, whose 
accuracy and reliability are documented and widely recognized, will be used to calculate the required 
natural periods and associated mode shapes, their use is not required. 
5.3.4 Modal base shear. A central feature of modal analysis is that the earthquake response is 
considered as a combination of the independent responses of the structure vibrating in each of its 
important modes. As the structure vibrates back and forth in a particular mode at the associated period, it 
experiences maximum values of base shear, story drifts, floor displacements, base (overturning) moments, 
etc. In this section, the base shear in the mth mode is specified as the product of the modal seismic 
coefficient Csm and the effective weight Wm for the mode. The coefficient Csm is determined for each 
mode from Eq. 5.3-3 using the spectral acceleration Sam at the associated period of the mode, Tm, in 
addition to the R, which is discussed elsewhere in the Commentary. An exception to this procedure 
occurs for higher modes of those structures that have periods shorter than 0.3 second and that are founded 
on soils of Site Class D, E, or F. For such modes, Eq. 5.3-4 is used. Equation 5.3-4 gives values ranging 
from 0.4SDS/R for very short periods to SDS/R for Tm = 0.3. Comparing these values to the limiting values 
of Cs of SDS/R for Site Class D, it is seen that the use of Eq. 5.3-4, when applicable, reduces the modal 
base shear. This is an approximation introduced in consideration of the conservatism embodied in using 
the spectral shape specified in Sec. 3.3.4. The spectral shape so defined is a conservative approximation 
to average spectra that are known to first ascend, level off, and then decay as period increases. The 
design spectrum defined in Sec. 3.3.4 is somewhat more conservative. For Site Classes A, B, and C, the 
ascending portion of the spectra is completed at or below periods of 0.1 to 0.2 second. On the other hand, 
for soft soils the ascent may not be completed until a larger period is reached. Equation 5.3-4 is then a 
replacement for the spectral shape for Site Classes D, E and F and short periods that is more consistent 
with spectra for measured accelerations. It was introduced because it was judged unnecessarily 
2 003 Commentary, Chapter 5 
78 
conservative to use Eq. 3.3-5 for modal analysis of structures assigned to Site Classes D, E, and F. The 
effective modal seismic weight given in Eq. 5.3-2 can be interpreted as specifying the portion of the 
weight of the structure that participates in the vibration of each mode. It is noted that Eq. 5.3-2 gives 
values of Wm that are independent of how the modes are normalized. 
The final equation of this section, Eq. 5.3-5, is to be used if a modal period exceeds 4 seconds. It can be 
seen that Eq. 5.3-5 and 5.3-3 coincide at Tm equal to 4 seconds so that the effect of using Eq. 5.3-5 is to 
provide a more rapid decrease in Csm as a function of the known characteristics of earthquake response 
spectra at intermediate and long periods. At intermediate periods, the average velocity spectrum of strong 
earthquake motions from large (magnitude 6.5 and larger) earthquakes is approximately constant, which 
implies that Csm should decrease as 1/Tm. For very long periods, the average displacement spectrum of 
strong earthquake motions becomes constant which implies that Csm, a form of acceleration spectrum, 
should decay as 1/Tm
2. The period at which the displacement response spectrum becomes constant 
depends on the size of the earthquake, being larger for great earthquakes, and a representative period of 4 
seconds was chosen to make the transition. 
5.3.5 Modal forces, deflections, and drifts. This section specifies the forces and displacements 
associated with each of the important modes of response. 
Modal forces at each level are given by Eq. 5.3-6 and 5.3-7 and are expressed in terms of the seismic 
weight assigned to the floor, the mode shape, and the modal base shear Vm. In applying the forces Fxm to 
the structure, the direction of the forces is controlled by the algebraic sign of fxm. Hence, the modal forces 
for the fundamental mode will all act in the same direction, but modal forces for the second and higher 
modes will change direction as one moves up the structure. The form of Eq. 5.3-6 is somewhat different 
from that usually employed in standard references and shows clearly the relation between the modal 
forces and the modal base shear. It, therefore, is a convenient form for calculation and highlights the 
similarity to Eq. 5.2-10 in the ELF procedure. 
The modal deflections at each level are specified by Eq. 5.3-8 and 5.3-9. These are the displacements 
caused by the modal forces Fxm considered as static forces and are representative of the maximum 
amplitudes of modal response for the essentially elastic motions envisioned within the concept of the 
seismic response modification coefficient R. If the mode under consideration dominates the earthquake 
response, the modal deflection under the strongest motion contemplated by the Provisions can be 
estimated by multiplying by the deflection amplification factor Cd. It should be noted that dxm is 
proportional to fxm (this can be shown with algebraic substitution for Fxm in Eq. 5.3-9) and will therefore 
change direction up and down the structure for the higher modes. 
5.3.6 Modal story shears and moments. This section merely specifies that the forces of Eq. 5.3-6 
should be used to calculate the shears and moments for each mode under consideration. In essence, the 
forces from Eq. 5.3-6 are applied to each mass, and linear static methods are used to calculate story shears 
and story overturning moments. The base shear that results from the calculation should agree with 
computed using Eq. 5.3-1. 
5.3.7 Design values. This section specifies the manner in which the values of story shear, moment, and 
drift and the deflection at each level are to be combined. The method used, in which the design value is 
the square root of the sum of the squares of the modal quantities, was selected for its simplicity and its 
wide familiarity (Clough and Penzien, 1975; Newmark and Rosenblueth, 1971; Wiegel, 1970). In 
general, it gives satisfactory results, but it is not always a conservative predictor of the earthquake 
response inasmuch as more adverse combinations of modal quantities than are given by this method of 
combination can occur. The most common instance where combination by use of the square root of the 
sum of the squares is unconservative occurs when two modes have very nearly the same natural period. 
In this case, the responses are highly correlated and the designer should consider combining the modal 
quantities more conservatively (Newmark and Rosenblueth, 1971). The complete quadratic combination 
(CQC) technique provides somewhat better results than the square-root-of-the-sum-of-the-squares method 
for the case of closely spaced modes. 
 Structural Analysis Procedures 
79 
This section also limits the reduction of base shear that can be achieved by modal analysis compared to 
use of the ELF procedure. Some reduction, where it occurs, is thought to be justified because the modal 
analysis gives a somewhat more accurate representation of the earthquake response. Some limit to the 
reduction permitted as a result of the calculation of longer natural periods is necessary because the actual 
periods of vibration may not be as long, even at moderately large amplitudes of motion, due to the 
stiffening effects of structural elements not a part of the seismic-force-resisting system and of 
nonstructural components. The limit is imposed by comparison to 85 percent of the base shear value 
computed using the ELF procedure. Where modal analysis predicts response quantities corresponding to 
a total base shear less than 85 percent of that which is computed using the ELF procedure, all response 
results must be scaled up to that level. Where modal analysis predicts response quantities in excess of 
those predicted by the ELF procedure, this is likely the result of significant higher mode participation and 
reduction to the values obtained from the ELF procedure is not permitted. 
5.3.8 Horizontal shear distribution. This section requires that the design story shears calculated in Sec. 
5.3.6 and the torsional moments prescribed in Sec. 5.2.4 be distributed to the vertical elements of the 
seismic resisting system as specified in Sec. 5.2.4 and as elaborated on in the corresponding section of 
this commentary. 
5.3.9 Foundation overturning. Because story moments are calculated mode by mode (properly 
recognizing that the direction of forces Fxm is controlled by the algebraic sign of fxm) and then combined to 
obtain the design values of story moments, there is no reason for reducing these design moments. This is 
in contrast with reductions permitted in overturning moments calculated from equivalent lateral forces in 
the analysis procedures of Sec. 5.2 (see Sec. 5.2.5 of this commentary). However, in the design of the 
foundation, the overturning moment calculated at the foundation-soil interface may be reduced by 10 
percent for the reasons mentioned in Sec. 5.2.5 of this commentary. 
5.3.10 P-delta effects. Sec. 5.2.6 of this commentary applies to this section. In addition, to obtain the 
story drifts when using the modal analysis procedure of Sec. 5.3, the story drift for each mode should be 
determined independently for each story. The story drift should not be determined from the differential of 
combined lateral structural deflections since this latter procedure will tend to mask the higher mode 
effects in longer period structures. 
5.4 LINEAR RESPONSE HISTORY PROCEDURE 
Linear response history analysis, also commonly known as time history analysis, is a numerically 
involved technique in which the response of a structural model to a specific earthquake ground motion 
accelerogram is determined through a process of numerical integration of the equations of motion. The 
ground shaking accelerogram, or record, is digitized into a series of small time steps, typically on the 
order of 1/100th of a second or smaller. Starting at the initial time step, a finite difference solution, or 
other numerical integration algorithm is followed to allow the calculation of the displacements of each 
node in the model and the forces in each element of model for each time step of the record. For even 
small structural models, this requires thousands of calculations and produces tens of thousands of data 
points. Clearly, such a calculation procedure can be performed only with the aid of high speed 
computers. However, even with the use of such computers, which are now commonly available, 
interpretation of the voluminous data that results from such analysis is tedious. 
The principal advantages of response history analysis, as opposed to response spectrum analysis, is that 
response history analysis provides a time dependent history of the response of the structure to a specific 
ground motion, allowing calculation of path dependent effects such as damping and also providing 
information on the stress and deformation state of the structure throughout the period of response. A 
response spectrum analysis, however, indicates only the maximum response quantities and does not 
indicate when during the period of response these occur, or how response of different portions of the 
structure is phased relative to that of other portions. Response history analyses are highly dependent on 
the characteristics of the individual ground shaking records and subtle changes in these records can lead 
to significant differences with regard to the predicted response of the structure. This is why, when 
response history analyses are used in the design process, it is necessary to run a suite of ground motion 
2 003 Commentary, Chapter 5 
80 
records. The use of multiple records in the analyses allows observation of the difference in response, 
resulting from differences in record characteristics. As a minimum, the Provisions require that suites of 
ground motions include at least three different records. However, suites containing larger numbers of 
records are preferable, since when more records are run, it is more likely that the differing response 
possibilities for different ground motion characteristics are observed. In order to encourage the use of 
larger suites, the Provisions require that when a suite contains fewer than seven records, the maximum 
values of the predicted response parameters be used as the design values. When seven or more records 
are used, then mean values of the response parameters may be used. This can lead to a substantial 
reduction in design forces and displacements and typically will justify the use of larger suites of records. 
Where possible, ground motion records should be scaled from actual recorded earthquake ground motions 
with characteristics (earthquake magnitude, distance from causative fault, and site soil conditions) similar 
to those which control the design earthquake for the site. Since only a limited number of actual 
recordings are available for such purposes, the use of synthetic records is permitted and may often be 
required. 
The extra complexity and cost inherent in the use of response history analysis rather than modal response 
spectrum analysis is seldom justified. As a result this procedure is rarely used in the design process. One 
exception is for the design of structures with energy dissipation systems comprising linear viscous 
dampers. Linear response history analysis can be used to predict the response of structures with such 
systems, while modal response spectrum analysis cannot. 
5.5 NONLINEAR RESPONSE HISTORY PROCEDURE 
This method of analysis is very similar to linear response history analysis, described in Sec. 5.4, except 
that the mathematical model is formulated in such a way that the stiffness and even connectivity of the 
elements can be directly modified based on the deformation state of the structure. This permits the effects 
of element yielding, buckling, and other nonlinear behavior on structural response to be directly 
accounted for in the analysis. It also permits the evaluation of such nonlinear behaviors as foundation 
rocking, opening and closing of gaps, and nonlinear viscous and hysteric damping. Potentially, this 
ability to directly account for these various nonlinearities can permit nonlinear response history analysis 
to provide very accurate evaluations of the response of the structure to strong ground motion. However, 
this accuracy can seldom be achieved in practice. This is partially because currently available nonlinear 
models for different elements can only approximate the behavior of real structural elements. Another 
limit on the accuracy of this approach is the fact that minor deviations in ground motion, such as those 
described in Sec. 5.4, or even in element hysteretic behavior, can result in significant differences in 
predicted response. For these reasons, when nonlinear response history analysis is used in the design 
process, suites of ground motion time histories must be considered, as described in Sec. 5.4. It may also 
be appropriate to perform sensitivity studies, in which the assumed hysteretic properties of elements are 
allowed to vary, within expected bounds, to allow evaluation of the effects of such uncertainties on 
predicted response. 
Application of nonlinear response history analysis to even the simplest structures requires large, high 
speed computers and complex computer software that has been specifically developed for this purpose. 
Several software packages have been in use for this purpose in universities for a number of years. These 
include the DRAIN family of programs and also the IDARC and IDARST family of programs. However, 
these programs have largely been viewed as experimental and are not generally accompanied by the same 
level of documentation and quality assurance typically found with commercially available software 
packages typically used in design offices. Although commercial software capable of performing 
nonlinear response history analyses has been available for several years, the use of these packages has 
generally been limited to complex aerospace, mechanical, and industrial applications. 
As a result of this, nonlinear response history analysis has mostly been used as a research (rather than 
design) tool until very recently. With the increasing adoption of base isolation and energy dissipation 
technologies in the structural design process, however, the need to apply this analysis technique in the 
design office has increased, creating a demand for more commercially available software. In response to 
 Structural Analysis Procedures 
81 
this demand, several vendors of commercial structural analysis software have modified their analysis 
programs to include limited nonlinear capability including the ability to model base isolation bearings, 
viscous dampers, and friction dampers. Some of these programs also have a limited library of other 
nonlinear elements including beam and truss elements. Such software provides the design office with the 
ability to begin to practically implement nonlinear response history analysis on design projects. 
However, such software is still limited, and it is expected that it will be some years before design offices 
can routinely expect to utilize this technique in the design of complex structures. 
5.5.3.1 Member strength. Nonlinear response history analysis is primarily a deformation-based 
procedure, in which the amount of nonlinear deformation imposed on elements by response to earthquake 
ground shaking is predicted. As a result, when this analysis method is employed, there is no general need 
to evaluate the strength demand (forces) imposed on individual elements of the structure. Instead, the 
adequacy of the individual elements to withstand the imposed deformation demands is directly evaluated, 
under the requirements of Sec. 5.5.3.2. The exception to this is the requirement to evaluate brittle 
elements, the failure of which could result in structural collapse, for the forces predicted by the analysis. 
These elements are identified in the Provisions through the requirement that they be evaluated for 
earthquake forces using the seismic effects defined in Sec. 4.2.2.2. That section requires that forces 
predicted by elastic analysis be amplified by a factor, O0
, to account in an approximate manner for the 
actual maximum force that can be delivered to the element, considering the inelastic behavior of the 
structure. Since nonlinear response history analysis does not use a response modification factor, as do 
elastic analysis approaches, and directly accounts for inelastic structural behavior, there is no need to 
further increase the forces by this factor. Instead the forces predicted by the analysis are used directly in 
the evaluation of the elements for adequacy under Sec. 4.2.2.2. 
5.5.4 Design review. The provisions for design using linear methods of analysis including the 
equivalent lateral force technique of Sec. 5.2 and the modal response spectrum analysis technique of Sec. 
5.3, are highly prescriptive. They limit the modeling assumptions that can be employed as well as the 
minimum strength and stiffness the structure must possess. Further, the methods used in linear analysis 
have become standardized in practice such that it is unlikely that different designers using the same 
technique to analyze the same structure will produce substantially different results. However, when 
nonlinear analytical methods are employed to predict the structures strength and its deformation under 
load, many of these prescriptive provisions are no longer applicable. Further, as these methods are 
currently not widely employed by the profession, the standardization that has occurred for linear methods 
of analysis has not yet been developed for these techniques. As a result analysis has not yet been 
developed for these techniques, and the designer using such methods must employ a significant amount of 
independent judgment in developing appropriate analytical models, performing the analysis, and 
interpreting the results to confirm the adequacy of a design. Since relatively minor changes in the 
assumptions used in performing a nonlinear structural analysis can significantly affect the results obtained 
from such an analysis, it is imperative that the assumptions used be appropriate. The Provisions require 
that designs employing nonlinear analysis methods be subjected to independent design review in order to 
provide a level of assurance that the independent judgment applied by the designer when using these 
methods is appropriate and compatible with that which would be made by other competent practitioners. 
5.6 SOIL-STRUCTURE INTERACTION EFFECTS 
5.6.1 General 
Statement of the problem. Fundamental to the design requirements presented in Sec. 5.2 and 5.3 is the 
assumption that the motion experienced by the base of a structure during an earthquake is the same as the 
free-field ground motion, a term that refers to the motion that would occur at the level of the foundation 
2 003 Commentary, Chapter 5 
82 
if no structure was present. This assumption implies that the foundation-soil system underlying the 
structure is rigid and, hence, represents a fixed-base condition. Strictly speaking, this assumption never 
holds in practice. For structures supported on a deformable soil, the foundation motion generally is 
different from the free-field motion and may include an important rocking component in addition to a 
lateral or translational component. The rocking component, and soil-structure interaction effects in 
general, tend to be most significant for laterally stiff structures such as buildings with shear walls, 
particularly those located on soft soils. For convenience, in what follows the response of a structure 
supported on a deformable foundation-soil system will be denoted as the flexible-base response. 
A flexibly supported structure also differs from a rigidly supported structure in that a substantial part of 
its vibrational energy may be dissipated into the supporting medium by radiation of waves and by 
hysteretic action in the soil. The importance of the latter factor increases with increasing intensity of 
ground-shaking. There is, of course, no counterpart of this effect of energy dissipation in a rigidly 
supported structure. 
The effects of soil-structure interaction accounted for in Sec. 5.6 represent the difference in the flexiblebase 
and fixed-base responses of the structure. This difference depends on the properties of the structure 
and the supporting medium as well as the characteristics of the free-field ground motion. 
The interaction effects accounted for in Sec. 5.6 should not be confused with site effects, which refer to 
the fact that the characteristics of the free-field ground motion induced by a dynamic event at a given site 
are functions of the properties and geological features of the subsurface soil and rock. The interaction 
effects, on the other hand, refer to the fact that the dynamic response of a structure built on that site 
depends, in addition, on the interrelationship of the structural characteristics and the properties of the 
local underlying soil deposits. The site effects are reflected in the values of the seismic coefficients 
employed in Sec. 5.2 and 5.3 and are accounted for only implicitly in Sec. 5.6. 
Possible approaches to the problem. Two different approaches may be used to assess the effects of 
soil-structure interaction. The first involves modifying the stipulated free-field design ground motion, 
evaluating the response of the given structure to the modified motion of the foundation, and solving 
simultaneously with additional equations that define the motion of the coupled system, whereas the 
second involves modifying the dynamic properties of the structure and evaluating the response of the 
modified structure to the prescribed free-field ground motion (Jennings and Bielak, 1973; Veletsos, 
1977). When properly implemented, both approaches lead to equivalent results. However, the second 
approach, involving the use of the free-field ground motion, is more convenient for design purposes and 
provides the basis of the requirements presented in Sec. 5.6. 
Characteristics of interaction. The interaction effects in the approach used here are expressed by an 
increase in the fundamental natural period of the structure and a change (usually an increase) in its 
effective damping. 
 Structural Analysis Procedures 
83 
The increase in period results from the flexibility of the foundation soil whereas the change in damping 
results mainly from the effects of energy dissipation in the soil due to radiation and material damping. 
These statements can be clarified by comparing the responses of rigidly and elastically supported systems 
subjected to a harmonic excitation of the base. 
Figure C5.6-1 Simple system investigated. 
Figure C5.2-3 Measured building period for concrete shear 
wall structures. 
2 003 Commentary, Chapter 5 
84 
Consider a linear structure of weight W, lateral stiffness k, and coefficient of viscous damping c (shown in 
Figure C5.6-1) and assume that it is supported by a foundation of weight Wo at the surface of a 
homogeneous, elastic halfspace. 
The foundation mat is idealized as a rigid circular plate of negligible thickness bonded to the supporting 
medium, and the columns of the structure are considered to be weightless and axially inextensible. Both 
the foundation weight and the weight of the structure are assumed to be uniformly distributed over 
circular areas of radius r. The base excitation is specified by the free-field motion of the ground surface. 
This is taken as a horizontally directed, simple harmonic motion with a period To and an acceleration 
amplitude am. 
The configuration of this system, which has three degrees of freedom when flexibly supported and a 
single degree of freedom when fixed at the base, is specified by the lateral displacement and rotation of 
the foundation, y and ., and by the displacement of the top of the structure, u, relative to its base. The 
system may be viewed either as the direct model of a one-story structural frame or, more generally, as a 
model of a multistory, multimode structure that responds as a single-degree-of-freedom system in its 
fixed-base condition. In the latter case, h must be interpreted as the distance from the base to the centroid 
of the inertia forces associated with the fundamental mode of vibration of the fixed-base structure and W, 
k, and c must be interpreted as its generalized or effective weight, stiffness, and damping coefficient, 
respectively. The relevant expressions for these quantities are given below. 
The solid lines in Figures C5.6-2 and C5.6-3 represent response spectra for the steady-state amplitude of 
the total shear in the columns of the system considered in Figure C5.6-1. Two different values of h/r and 
several different values of the relative flexibility parameter for the soil and the structure, fo
, are 
considered. The latter parameter is defined by the equation 0 
s
h 
v T 
d = in which h is the height of the 
structure as previously indicated, vs is the velocity of shear wave propagation in the halfspace, and T is the 
fixed-base natural period of the structure. A value of f = 0 corresponds to a rigidly supported structure. 
The results in Figures C5.6-2 and C5.6-3 are displayed in a dimensionless form, with the abscissa 
representing the ratio of the period of the excitation, To, to the fixed-base natural period of the system, T, 
and the ordinate representing the ratio of the amplitude of the actual base shear, V, to the amplitude of the 
base shear induced in an infinitely stiff, rigidly supported structure. 
The latter quantity is given by the product mam, in which m = W/g, g is the acceleration due to gravity, 
and am is the acceleration amplitude of the free-field ground motion. The inclined scales on the left 
represent the deformation amplitude of the superstructure, u, normalized with respect to the displacement 
amplitude of the free-field ground motion 
2 
0
2 . 
4
m 
m 
d a T 
p 
= 
 Structural Analysis Procedures 
85 
The damping of the structure in its fixed-base condition, , is considered to be 2 percent of the critical 
value, and the additional parameters needed to characterize completely these solutions are identified in 
Veletsos and Meek (1974), from which these figures have been reproduced. 
Comparison of the results presented in these figures reveals that the effects of soil-structure interaction 
are most strikingly reflected in a shift of the peak of the response spectrum to the right and a change in the 
magnitude of the peak. These changes, which are particularly prominent for taller structures and more 
flexible soils (increasing values of f0
), can conveniently be expressed by an increase in the natural period 
of the system over its fixed-base value and by a change in its damping factor. 
Also shown in these figures in dotted lines are response spectra for single-degree-of-freedom (SDF) 
oscillators, the natural period and damping of which have been adjusted so that the absolute maximum 
(resonant) value of the base shear and the associated period are in each case identical to those of the 
actual interacting systems. The base motion for the replacement oscillator is considered to be the same as 
the free-field ground motion. With the properties of the replacement SDF oscillator determined in this 
manner, it is important to note that the response spectra for the actual and the replacement systems are in 
excellent agreement over wide ranges of the exciting period on both sides of the resonant peak. 
Figure C5.6-2 Response spectra for systems with h/r = 1 
(Veletsos and Meek, 1974). 
2 003 Commentary, Chapter 5 
86 
In the context of Fourier analysis, an earthquake motion may be viewed as the result of superposition of 
harmonic motions of different periods and amplitudes. Inasmuch as the components of the excitation 
with periods close to the resonant period are likely to be the dominant contributors to the response, the 
maximum responses of the actual system and of the replacement oscillator can be expected to be in 
satisfactory agreement for earthquake ground motions as well. This expectation has been confirmed by 
the results of comprehensive comparative studies (Veletsos, 1977; Veletsos and Meek, 1974; Veletsos 
and Nair, 1975; Jennings and Bielak, 1973). 
It follows that, to the degree of approximation involved in the representation of the actual system by the 
replacement SDF oscillator, the effects of interaction on maximum response may be expressed by an 
increase in the fundamental natural period of the fixed-base system and by a change in its damping value. 
In the following sections, the natural period of replacement oscillator is denoted by T% and the associated 
damping factor by %. These quantities will also be referred to as the effective natural period and the 
effective damping factor of the interacting system. The relationships between T% and T and between % and 
 are considered in Sec. 5.6.2.1.1 and 5.6.2.1.2. 
Figure C5.6-3 Response spectra for systems with 
h/r = 5 (Veletsos and Meek, 1974). 
 Structural Analysis Procedures 
87 
Basis of provisions and assumptions. Current knowledge of the effects of soil-structure interactions is 
derived mainly from studies of systems of the type referred to above in which the foundation is idealized 
as a rigid mat. For foundations of this type, both surface-supported and embedded structures resting on 
uniform as well as layered soil deposits have been investigated (Bielak, 1975; Chopra and Gutierrez, 
1974; Jennings and Bielak, 1973; Liu and Fagel, 1971; Parmelee et al., 1969; Roesset et al., 1973; 
Veletsos, 1977; Veletsos and Meek, 1974; Veletsos and Nair, 1975). However, the results of such studies 
may be of limited applicability for foundation systems consisting of individual spread footings or deep 
foundations (piles or drilled shafts) not interconnected with grade beams or a mat. The requirements 
presented in Sec. 5.6 for the latter cases represent the best interpretation and judgment of the developers 
of the requirements regarding the current state of knowledge. 
Fundamental to these requirements is the assumption that the structure and the underlying soil are bonded 
and remain so throughout the period of ground-shaking. It is further assumed that there is no soil 
instability or large foundation settlements. The design of the foundation in a manner to ensure 
satisfactory soil performance (for example, to avoid soil instability and settlement associated with the 
compaction and liquefaction of loose granular soils), is beyond the scope of Sec. 5.6. Finally, no account 
is taken of the interaction effects among neighboring structures. 
Nature of interaction effects. Depending on the characteristics of the structure and the ground motion 
under consideration, soil-structure interaction may increase, decrease, or have no effect on the magnitudes 
of the maximum forces induced in the structure itself (Bielak, 1975; Jennings and Bielak, 1973; Veletsos, 
1977; Veletsos and Meek, 1974; Veletsos and Nair, 1975). However, for the conditions stipulated in the 
development of the requirements for rigidly supported structures presented in Sec. 5.2 and 5.3, soilstructure 
interaction will reduce the design values of the base shear and moment from the levels 
applicable to a fixed-base condition. These forces therefore can be evaluated conservatively without the 
adjustments recommended in Sec. 5.6. 
Because of the influence of foundation rocking, however, the horizontal displacements relative to the base 
of the elastically supported structure may be larger than those of the corresponding fixed-base structure, 
and this may increase both the required spacing between structures and the secondary design forces 
associated with the P-delta effects. Such increases generally are small for frame structures, but can be 
significant for shear wall structures. 
Scope. Two procedures are used to incorporate effects of the soil-structure interaction. The first is an 
extension of the equivalent lateral force procedure presented in Sec. 5.2 and involves the use of 
equivalent lateral static forces. The second is an extension of the simplified modal analysis procedure 
presented in Sec. 5.3. In the latter approach, the earthquake-induced effects are expressed as a linear 
combination of terms, the number of which is equal to the number of stories involved. Other more 
complex procedures also may be used, and these are outlined briefly at the end of this commentary on 
Sec. 5.6. However, it is believed that the more involved procedures are justified only for unusual 
structures and when the results of the specified simpler approaches have revealed that the interaction 
effects are indeed of definite consequence in the design. 
5.6.2 Equivalent lateral force procedure. This procedure is similar to that used in the older SEAOC 
recommendations except that it incorporates several improvements (see Sec. 5.2 of this commentary). In 
effect, the procedure considers the response of the structure in its fundamental mode of vibration and 
accounts for the contributions of the higher modes implicitly through the choice of the effective weight of 
the structure and the vertical distribution of the lateral forces. The effects of soil-structure interaction are 
accounted for on the assumption that they influence only the contribution of the fundamental mode of 
vibration. For structures, this assumption has been found to be adequate (Bielak, 1976; Jennings and 
Bielak, 1973; Veletsos, 1977). 
5.6.2.1 Base shear. With the effects of soil-structure interaction neglected, the base shear is defined by 
Eq. 5.2-1, V = CsW, in which W is the total seismic weight (as specified in Sec. 5.2.1) and Cs is the 
dimensionless seismic response coefficient (as defined in Sec. 5.2.1.1). This term depends on the level of 
seismic hazard under consideration, the properties of the site, and the characteristics of the structure itself. 
2 003 Commentary, Chapter 5 
88 
The latter characteristics include the rigidly supported fundamental natural period of the structure, T, the 
associated damping factor, , and the degree of permissible inelastic deformation. The damping factor 
does not appear explicitly in Sec. 5.2.1.1 because a constant value of  = 0.05 has been used for all 
structures for which the interaction effects are negligible. The degree of permissible inelastic action is 
reflected in the choice of the reduction factor, R. It is convenient to rewrite Eq. 5.2-1 in the form: 
V=Cs(T,)W+Cs(T,)[W-W] (C5.6-1) 
where W represents the generalized or effective weight of the structure when vibrating in its fundamental 
natural mode. The terms in parentheses are used to emphasize the fact that Cs depends upon both T and . 
The relationship between W and W is given below. The first term on the right side of Eq. C5.6-1 
approximates the contribution of the fundamental mode of vibration whereas the second term 
approximates the contributions of the higher natural modes. Inasmuch as soil-structure interaction may 
be considered to affect only the contribution of the fundamental mode and inasmuch as this effect can be 
expressed by changes in the fundamental natural period and the associated damping of the system, the 
base shear for the interacting system, V%,may be stated (in a form analogous to Eq. C5.6-1) as follows: 
V%=Cs(T%,%)W+Cs(T,)[W-W] (C5.6-2) 
The value of Cs in the first part of this equation should be evaluated for the natural period and damping of 
the elastically supported system, T% and %, respectively, and the value of Cs in the second term part should 
be evaluated for the corresponding quantities of the rigidly supported system, T and . 
Before proceeding with the evaluation of the coefficients Cs in Eq. C5.6-2, it is desirable to rewrite this 
formula in the same form as Eq. 5.6-1. Making use of Eq. 5.2-1 and rearranging terms, the following 
expression for the reduction in the base shear is obtained: 
.V=..Cs(T,)-Cs(T%,%)..W (C5.6-3) 
Within the ranges of natural period and damping that are of interest in studies of structural response, the 
values of Cs corresponding to two different damping values but the same natural period (T), are related 
approximately as follows: 
0.4 
Cs(T, ) Cs(T, ) 
 
  
 
% % % 
% 
. . 
= . . 
. . 
(C5.6-4) 
This expression, which appears to have been first proposed in Arias and Husid (1962), is in good 
agreement with the results of studies of earthquake response spectra for systems having different damping 
values (Newmark et al., 1973). 
Substitution of Eq. C5.6-4 in Eq. C5.6-3 leads to: 
0.4 
V Cs(T, ) Cs(T, ) W 
 
  
 
% 
% 
. . .. 
. =. - . .. 
.. . ... 
(C5.6-5) 
where both values of Cs are now for the damping factor of the rigidly supported system and may be 
evaluated from Eq. 5.2-2 and 5.2-3. If the terms corresponding to the periods T and T% are denoted more 
simply as Cs and C%s , respectively, and if the damping factor  is taken as 0.05, Eq. C5.6-5 reduces to 
Eq. 5.6-2. 
Note that C%s in Eq. 5.6-2 is smaller than or equal to Cs because Eq. 5.2-3 is a nonincreasing function of 
the natural period and T% is greater than or equal to T. Furthermore, since the minimum value of % is 
taken as % = = 0.05 (see statement following Eq. 5.6-10), the shear reduction .V is a non-negative 
 Structural Analysis Procedures 
89 
quantity. It follows that the design value of the base shear for the elastically supported structure cannot 
be greater than that for the associated rigid-base structure. 
The effective weight of the structure, W, is defined by Eq. 5.3-2, in which fim should be interpreted as the 
displacement amplitude of the ith floor when the structure is vibrating in its fixed-base fundamental 
natural mode. It should be clear that the ratio W/W depends on the detailed characteristics of the 
structure. A constant value of W= 0.7 Wis recommended in the interest of simplicity and because it is a 
good approximation for typical structures. As an example, it is noted that for a tall structure for which the 
weight is uniformly distributed along the height and for which the fundamental natural mode increases 
linearly from the base to the top, the exact value of W= 0.7 W. Naturally, when the full weight of the 
structure is concentrated at a single level, W should be taken equal to W. 
The maximum permissible reduction in base shear due to the effects of soil-structure interaction is set at 
30 percent of the value calculated for a rigid-base condition. It is expected, however, that this limit will 
control only infrequently and that the calculated reduction, in most cases, will be less. 
5.6.2.1.1 Effective building period. Equation 5.6-3 for the effective natural period of the elastically 
supported structure, T%, is determined from analyses in which the superstructure is presumed to respond in 
its fixed-base fundamental mode and the foundation weight is considered to be negligible in comparison 
to the weight of the superstructure (Jennings and Bielak, 1973; Veletsos and Meek, 1974). The first term 
under the radical represents the period of the fixed-base structure. The first portion of the second term 
represents the contribution to T% of the translational flexibility of the foundation, and the last portion 
represents the contribution of the corresponding rocking flexibility. The quantities k and h represent, 
respectively, the effective stiffness and effective height of the structure, and Ky and K. represent the 
translational and rocking stiffnesses of the foundation. 
Equation 5.6-4 for the structural stiffness, k, is deduced from the well known expression for the natural 
period of the fixed-base system: 
T 2 1 W 
g k 
p 
. .. . 
= . .. . 
. .. . 
(C5.6-6) 
The effective height, h, is defined by Eq. 5.6-13, in which fi1 has the same meaning as the quantity fim in 
Eq. 5.3-2 when m = 1. In the interest of simplicity and consistency with the approximation used in the 
definition of W, however, a constant value of h= 0.7hn is recommended where hn is the total height of 
the structure. This value represents a good approximation for typical structures. As an example, it is 
noted that for tall structures for which the fundamental natural mode increases linearly with height, the 
exact value of h is 2/3hn. Naturally, when the gravity load of the structure is effectively concentrated at a 
single level, hn must be taken as equal to the distance from the base to the level of weight concentration. 
Foundation stiffnesses depend on the geometry of the foundation-soil contact area, the properties of the 
soil beneath the foundation, and the characteristics of the foundation motion. Most of the available 
information on this subject is derived from analytical studies of the response of harmonically excited rigid 
circular foundations, and it is desirable to begin with a brief review of these results. 
For circular mat foundations supported at the surface of a homogeneous halfspace, stiffnesses Ky and K. 
are given by: 
8
2 
y 
Ky Gr 
v 
. a . 
=.. - .. 
(C5.6-7) 
and 
2 003 Commentary, Chapter 5 
90 
( ) 
8 3 
3 1 
K v Gr 
. 
. 
. a . 
=. . 
.. - .. 
(C5.6-8) 
where r is the radius of the foundation; G is the shear modulus of the halfspace; v is its Poissons ratio; 
and ay and a. are dimensionless coefficients that depend on the period of the excitation, the dimensions of 
the foundation, and the properties of the supporting medium (Luco, 1974; Veletsos and Verbic, 1974; 
Veletsos and Wei, 1971). The shear modulus is related to the shear wave velocity, vs, by the formula: 
2
G vs 
g 
. 
= (C5.6-9) 
in which . is the unit weight of the material. The values of G, vs, and v should be interpreted as average 
values for the region of the soil that is affected by the forces acting on the foundation and should 
correspond to the conditions developed during the design earthquake. The evaluation of these quantities 
is considered further in subsequent sections. For statically loaded foundations, the stiffness coefficients 
ay and a. are unity, and Eq. C5.6-7 and C5.6-8 reduce to: 
8
y 2
K Gr
v 
= 
- 
(C5.6-10) 
and 
8 3 
3(1 ) 
K Gr
v . = 
- 
(C5.6-11) 
Studies of the interaction effects in structure-soil systems have shown that, within the ranges of 
parameters of interest for structures subjected to earthquakes, the results are insensitive to the 
period-dependency of ay and that it is sufficiently accurate for practical purposes to use the static stiffness 
Ky, defined by Eq. C5.6-10. However, the dynamic modifier for rocking a. can significantly affect the 
response of building structures. In the absence of more detailed analyses, for ordinary building structures 
with an embedment ratio d/r < 0.5, the factor a. can be estimated as follows: 
R/vsT 
a. 
<0.05 
1.0 
0.15 
0.85 
0.35 
0.7 
0.5 
0.6 
where d equals depth of embedment and r can be taken as rm defined in Eq. 5.6-8. 
The above values were derived from the solution for a. by Veletsos and Verbic (1973). In this solution a. 
is a function of T%. To relate a. to T, a correction for period lengthening (T%/T) was made assuming 
h r .. 0.5 to1.0 and Poisson=s ratio v = 0.4. 
Foundation embedment has the effect of increasing the stiffnesses Ky and K.. For embedded foundations 
for which there is positive contact between the side walls and the surrounding soil, Ky and K. may be 
determined from the following approximate formulas: 
8 1 2 
y 2 3 
K Gr d 
v r 
=... - ......+... ...... ...... 
(C5.6-12) 
and 
 Structural Analysis Procedures 
91 
8 3 1 2 
3(1 ) 
K Gr d 
v r 
. 
. 
=... -a ......+ ... ...... 
(C5.6-13) 
in which d is the depth of embedment. These formulas are based on finite element solutions (Kausel, 
1974). 
Both analyses and available test data (Erden, 1974) indicate that the effects of foundation embedment are 
sensitive to the condition of the backfill and that judgment must be exercised in using Eq. C5.6-12 and 
C5.6-13. For example, if a structure is embedded in such a way that there is no positive contact between 
the soil and the walls of the structure, or when any existing contact cannot reasonably be expected to 
remain effective during the stipulated design ground motion, stiffnesses Ky and K. should be determined 
from the formulas for surface-supported foundations. More generally, the quantity d in Eq. C5.6-12 and 
C5.6-13 should be interpreted as the effective depth of foundation embedment for the conditions that 
would prevail during the design earthquake. 
The formulas for Ky and K. presented above are strictly valid only for foundations supported on 
reasonably uniform soil deposits. When the foundation rests on a surface stratum of soil underlain by a 
stiffer deposit with a shear wave velocity (vs) more than twice that of the surface layer (Wallace et al., 
1999), Ky and K. may be determined from the following two generalized formulas in which G is the shear 
modulus of the soft soil and Ds is the total depth of the stratum. First, using Eq. C5.6-12: 
8 12 11 15 
y 2 3 2 4 
s s 
K Gr d r d 
v r D D 
=... - ......+... ...... ..........+... ...... ...........+... ...... ....... 
(C5.6-14) 
Second, using Eq. C5.6-13: 
8 3 1 2 1 1 1 0.7 
3(1 ) 6 s s 
K Gr d r d 
v r D D 
. 
. 
=... -a ......+ ... ..........+... ...... ...........+ ... ....... 
(C5.6-15) 
These formulas are based on analyses of a stratum supported on a rigid base (Elsabee et al., 1977; Kausel 
and Roesset, 1975) and apply for r/Ds < 0.5 and d/r < 1. 
The information for circular foundations presented above may be applied to mat foundations of arbitrary 
shapes provided the following changes are made: 
1. The radius r in the expressions for Ky is replaced by ra (Eq. 5.6-7), which represents the radius of a 
disk that has the area, Ao, of the actual foundation. 
2. The radius r in the expressions for K. is replaced by rm (Eq. 5.6-8), which represents the radius of a 
disk that has the moment of inertia, Io, of the actual foundation. 
For footing foundations, stiffnesses Ky and K. are computed by summing the contributions of the 
individual footings. If it is assumed that the foundation behaves as a rigid body and that the individual 
footings are widely spaced so that they act as independent units, the following formulas are obtained: 
Ky= Skyi (C5.6-16) 
and 
2 
K. = Skxiyi + Sk. i (C5.6-17) 
The quantity kyi represents the horizontal stiffness of the ith footing; kxi and k.i represent, respectively, the 
corresponding vertical and rocking stiffnesses; and yi represents the normal distance from the centroid of 
the ith footing to the rocking axis of the foundation. The summations are considered to extend over all 
footings. The contribution to K. of the rocking stiffnesses of the individual footings, k.i, generally is 
small and may be neglected. 
The stiffnesses kyi, kxi, and k.i are defined by the formulas: 
2 003 Commentary, Chapter 5 
92 
8 1 2 
2 3 
i ai 
yi 
k G r d 
v r 
=... - ......+ ... 
(C5.6-18) 
4 1 0.4 
1 
i ai 
yi 
k G r d 
v r 
=... - ......+ ... 
(C5.6-19) 
and 
8 3 1 2 
3(1 ) 
i mi 
i 
k G r d 
v r . 
=... - ......+ ... 
(C5.6-20) 
in which di is the depth of effective embedment for the ith footing; Gi is the shear modulus of the soil 
beneath the ith footing; rai = Aoi /p is the radius of a circular footing that has the area of the ith footing, 
Aoi; and rmi equals 44Ioi/p the radius of a circular footing, the moment of inertia of which about a 
horizontal centroidal axis is equal to that of the ith footing, Ioi, in the direction in which the response is 
being evaluated. 
For surface-supported footings and for embedded footings for which the side wall contact with the soil 
cannot be considered to be effective during the stipulated design ground motion, di in these formulas 
should be taken as zero. Furthermore, the values of Gi should be consistent with the stress levels 
expected under the footings and should be evaluated with due regard for the effects of the dead loads 
involved. This matter is considered further in subsequent sections. For closely spaced footings, 
consideration of the coupling effects among footings will reduce the computed value of the overall 
foundation stiffness. This reduction, in turn, will increase the fundamental natural period of the system, 
T%, and increase the value of .V, the amount by which the base shear is reduced due to soil-structure 
interaction. It follows that the use of Eq. C5.6-16 and 5.6-17 will err on the conservative side in this case. 
The degree of conservatism involved, however, will partly be compensated by the presence of a basement 
slab that, even when it is not tied to the structural frame, will increase the overall stiffness of the 
foundation. 
The values of Ky and K. for pile foundations can be computed in a manner analogous to that described in 
the preceding section by evaluating the horizontal, vertical, and rocking stiffnesses of the individual piles, 
kyi, kxi, and k.i, and by combining these stiffnesses in accordance with Eq. C5.6-16 and C5.6-17. 
The individual pile stiffnesses may be determined from field tests or analytically by treating each pile as a 
beam on an elastic subgrade. Numerous formulas are available in the literature (Tomlinson, 1994) that 
express these stiffnesses in terms of the modulus of the subgrade reaction and the properties of the pile 
itself. These stiffnesses sometimes are expressed in terms of the stiffness of an equivalent freestanding 
cantilever, the physical properties and cross-sectional dimensions of which are the same as those of the 
actual pile but the length of which is adjusted appropriately. The effective lengths of the equivalent 
cantilevers for horizontal motion and for rocking or bending motion are slightly different but are often 
assumed to be equal. On the other hand, the effective length in vertical motion is generally considerably 
greater. 
The soil properties of interest are the shear modulus, G, or the associated shear wave velocity, vs; the unit 
weight, .; and Poissons ratio, v. These quantities are likely to vary from point to point of a construction 
site, and it is necessary to use average values for the soil region that is affected by the forces acting on the 
foundation. The depth of significant influence is a function of the dimensions of the foundation base and 
of the direction of the motion involved. The effective depth may be considered to extend to about 0.75ra 
below the foundation base for horizontal motions, 2ra for vertical motions, and to about 0.75rm for 
rocking motion. For mat foundations, the effective depth is related to the total plan dimensions of the mat 
whereas for structures supported on widely spaced spread footings, it is related to the dimensions of the 
individual footings. For closely spaced footings, the effective depth may be determined by superposition 
of the pressure bulbs induced by the forces acting on the individual footings. 
 Structural Analysis Procedures 
93 
Since the stress-strain relations for soils are nonlinear, the values of G and vs also are functions of the 
strain levels involved. In the formulas presented above, G should be interpreted as the secant shear 
modulus corresponding to the significant strain level in the affected region of the foundation soil. The 
approximate relationship of this modulus to the modulus Go corresponding to small amplitude strains (of 
the order of 10-3 percent or less) is given in Table 5.6-1. The backgrounds of this relationship and of the 
corresponding relationship for vs/vso are identified below. 
The low amplitude value of the shear modulus, Go, can most conveniently be determined from the 
associated value of the shear wave velocity, vso, by use of Eq. C5.6-9. The latter value may be determined 
approximately from empirical relations or more accurately by means of field tests or laboratory tests. 
The quantities Go and vso depend on a large number of factors (Hardin, 1978), the most important of 
which are the void ratio, e, and the average confining pressure, s o . The value of the latter pressure at a 
given depth beneath a particular foundation may be expressed as the sum of two terms as follows: 
so=sos+sob (C5.6-21) 
in which s os represents the contribution of the weight of the soil and s ob represents the contribution of 
the superimposed weight of the structure and foundation. The first term is defined by the formula: 
1 2
3 
o 
os 
s K . x 
=.. + .. ' 
. . 
(C5.6-22) 
in which x is the depth of the soil below the ground surface, . ' is the average effective unit weight of the 
soil to the depth under consideration, and Ko is the coefficient of horizontal earth pressure at rest. For 
sands and gravel, Ko has a value of 0.5 to 0.6 whereas for soft clays, Ko  1.0. The pressures s ob 
developed by the weight of the structure can be estimated from the theory of elasticity (Poulos and Davis, 
1974). In contrast to s os which increases linearly with depth, the pressures s ob decrease with depth. As 
already noted, the value of vso should correspond to the average value of s o in the region of the soil that 
is affected by the forces acting on the foundation. 
For clean sands and gravels having e < 0.80, the low-amplitude shear wave velocity can be calculated 
approximately from the formula: 
0.25 
vso=c1(2.17-e)(s ) (C5.6-23) 
in which c1 equals 78.2 when s is in lb/ft2 and vso is in ft/sec; c1 equals 160.4 when s is in kg/cm2 and 
vso is in m/sec; and c1 equals 51.0 when s is in kN/m2 and vso is in m/sec. 
For angular-grained cohesionless soils (e > 0.6), the following empirical equation may be used: 
0.25 
vso=c2 (2.97-e)(s ) (C5.6-24) 
in which c2 equals 53.2 when s is in lb/ft2 and vso is in ft/sec; c2 equals 109.7 when s is in kg/cm2 and 
vso is in m/sec; and c2 equals 34.9 when s is in kN/m2 and vso is in m/sec. 
Equation C5.6-24 also may be used to obtain a first-order estimate of vso for normally consolidated 
cohesive soils. A crude estimate of the shear modulus, Go, for such soils may also be obtained from the 
relationship: 
Go= 1,000su (C5.6-25) 
in which su is the shearing strength of the soil as developed in an unconfined compression test. The 
coefficient 1,000 represents a typical value, which varied from 250 to about 2,500 for tests on different 
soils (Hara et al., 1974; Hardin and Drnevich, 1975). 
These empirical relations may be used to obtain preliminary, order-of-magnitude estimates. For more 
accurate evaluations, field measurements of vso should be made. Field evaluations of the variations of vso 
2 003 Commentary, Chapter 5 
94 
throughout the construction site can be carried out by standard seismic refraction methods, the downhole 
or cross-hole methods, suspension logging, or spectral analysis with surface waves. Kramer (1996) 
provides an overview of these testing procedures. The disadvantage of these methods is that vso is 
determined only for the stress conditions existing at the time of the test (usually s so ). The effect of the 
changes in the stress conditions caused by construction must be considered by use of Eq. C5.6-22, 
C5.6-23, and C5.6-24 to adjust the field measurement of vso to correspond to the prototype situations. 
The influence of large-amplitude shearing strains may be evaluated from laboratory tests or approximated 
through the use of Table 5.6-1. This matter is considered further in the next two sections. 
An increase in the shearing strain amplitude is associated with a reduction in the secant shear modulus, G, 
and the corresponding value of vs. Extensive laboratory tests (for example, Vucetic and Dobry, 1991; 
Seed et al., 1984) have established the magnitudes of the reductions in vs for both sands and clays as the 
shearing strain amplitude increases. 
The results of such tests form the basis for the information presented in Table 5.6-1. For each severity of 
anticipated ground-shaking, represented by the effective peak acceleration coefficients (taken as 0.4SDS) a 
representative value of shearing strain amplitude was developed. A conservative value of vs/vso that is 
appropriate to that strain amplitude then was established. It should be emphasized that the values in 
Table 5.6-1 are first order approximations. More precise evaluations would require the use of materialspecific 
shear modulus reduction curves and studies of wave propagation for the site to determine the 
magnitude of the soil strains induced. 
It is satisfactory to assume Poissons ratio for soils as: v = 0.33 for clean sands and gravels, v = 0.40 for 
stiff clays and cohesive soils, and v = 0.45 for soft clays. The use of an average value of v = 0.4 also will 
be adequate for practical purposes. 
Regarding an alternative approach, note that Eq. 5.6-5 for the period T% of structures supported on mat 
foundations was deduced from Eq. 5.6-3 by making use of Eq. C5.6-10 and C5.6-11, with Poissons ratio 
taken as v = 0.4 and with the radius r interpreted as ra in Eq. C5.6-10 and as rm in Eq. C5.6-11. For a 
nearly square foundation, for which ra  rm  r, Eq. 5.6-5 reduces to: 
2 
2 2 2 
1 25 1 1.12 
s
T T rh h 
v T . r 
a 
a 
% 
. .. . .. 
= + . ..+. .. 
. ... . ... 
(C5.6-26) 
The value of the relative weight parameter, a, is likely to be in the neighborhood of 0.15 for typical 
structures. 
5.6.2.1.2 Effective damping. Equation 5.6-9 for the overall damping factor of the elastically supported 
structure, %, was determined from analyses of the harmonic response at resonance of simple systems of 
the type considered in Figures C5.6-2 and 5.6-3. The result is an expression of the form (Bielak, 1975; 
Veletsos and Nair, 1975) of: 
3 
0.05 
0 
T
T 
%  
% 
= + 
. . 
. . 
. . 
(C5.6-27) 
in which o represents the contribution of the foundation damping, considered in greater detail in the 
following paragraphs, and the second term represents the contribution of the structural damping. The 
latter damping is assumed to be of the viscous type. Equation C5.6-27 corresponds to the value of 
 = 0.05 used in the development of the response spectra for rigidly supported systems employed in 
Sec. 5.2. 
The foundation damping factor, o, incorporates the effects of energy dissipation in the soil due to the 
following sources: the radiation of waves away from the foundation, known as radiation or geometric 
damping, and the hysteretic or inelastic action in the soil, also known as soil material damping. This 
 Structural Analysis Procedures 
95 
factor depends on the geometry of the foundation-soil contact area and on the properties of the structure 
and the underlying soil deposits. 
For mat foundations of circular plan that are supported at the surface of reasonably uniform soils deposits, 
the three most important parameters which affect the value of o are: the ratio (T% /T) of the fundamental 
natural periods of the elastically supported and the fixed-base structures, the ratio h r of the effective 
height of the structure to the radius of the foundation, and the damping capacity of the soil. The latter 
capacity is measured by the dimensionless ratio .Ws/Ws, in which .Ws is the area of the hysteresis loop in 
the stress-strain diagram for a soil specimen undergoing harmonic shearing deformation and Ws is the 
strain energy stored in a linearly elastic material subjected to the same maximum stress and strain (that is, 
the area of the triangle in the stress-strain diagram between the origin and the point of the maximum 
induced stress and strain). This ratio is a function of the magnitude of the imposed peak strain, increasing 
with increasing intensity of excitation or level of strain. 
The variation of o with T% T and h r is given in Figure 5.6-1 for two levels of excitation. The dashed 
lines, which are recommended for values of the effective peak ground acceleration (taken as 0.4SDS) equal 
to or less than 0.10, correspond to a value of .Ws/Ws  0.3, whereas the solid lines, which are 
recommended for values of effective peak ground acceleration equal to or greater than 0.20, correspond to 
a value of .Ws/Ws  1. These curves are based on the results of extensive parametric studies (Veletsos, 
1977; Veletsos and Meek, 1974; Veletsos and Nair, 1975) and represent average values. For the ranges 
of parameters that are of interest in practice, however, the dispersion of the results is small. 
For mat foundations of arbitrary shape, the quantity r in Figure 5.6-1 should be interpreted as a 
characteristic length that is related to the length of the foundation, Lo, in the direction in which the 
structure is being analyzed. For short, squatty structures for which h/Lo =0.5, the overall damping of 
the structure-foundation system is dominated by the translational action of the foundation, and it is 
reasonable to interpret r as ra, the radius of a disk that has the same area as that of the actual foundation 
(see Eq. 5.6-7). On the other hand, for structures with h/Lo =0.1,the interaction effects are dominated 
by the rocking motion of the foundation, and it is reasonable to define r as the radius rm of a disk whose 
static moment of inertia about a horizontal centroidal axis is the same as that of the actual foundation 
normal to the direction in which the structure is being analyzed (see Eq. 5.6-8). 
Subject to the qualifications noted in the following section, the curves in Figure 5.6-1 also may be used 
for embedded mat foundations and for foundations involving spread footings or piles. In the latter cases, 
the quantities Ao and Io in the expressions for the characteristic foundation length, r, should be interpreted 
as the area and the moment of inertia of the load-carrying foundation. 
In the evaluation of the overall damping of the structure-foundation system, no distinction has been made 
between surface-supported foundations and embedded foundations. Since the effect of embedment is to 
increase the damping capacity of the foundation (Bielak, 1975; Novak, 1974; Novak and Beredugo, 1972) 
and since such an increase is associated with a reduction in the magnitude of the forces induced in the 
structure, the use of the recommended requirements for embedded structures will err on the conservative 
side. 
There is one additional source of conservatism in the application of the recommended requirements to 
structures with embedded foundations. It results from the assumption that the free-field ground motion at 
the foundation level is independent of the depth of foundation embedment. Actually, there is evidence to 
the effect that the severity of the free-field excitation decreases with depth (Seed et al., 1977). This 
reduction is ignored both in Sec. 5.6 and in the requirements for rigidly supported structures presented in 
Sec. 5.2 and 5.3. 
Equations 5.6-9 and C5.6-28, in combination with the information presented in Figure 5.6-1, may lead to 
damping factors for the structure-soil system, %, that are smaller than the structural damping factor, . 
However, since the representative value of  = 0.05 used in the development of the design requirements 
2 003 Commentary, Chapter 5 
96 
for rigidly supported structures is based on the results of tests on actual structures, it reflects the damping 
of the full structure-soil system, not merely of the component contributed by the superstructure. Thus, the 
value of % determined from Eq. 5.6-9 should never be taken less than , and a minimum value of 
% = =0.05 has been imposed. The use of values of % >  is justified by the fact that the experimental 
values correspond to extremely small amplitude motions and do not reflect the effects of the higher soil 
damping capacities corresponding to the large soil strain levels associated with the design ground 
motions. The effects of the higher soil damping capacities are appropriately reflected in the values of o 
presented in Figure 5.6-1. 
There are, however, some exceptions. For foundations involving a soft soil stratum of reasonably 
uniform properties underlain by a much stiffer, rock-like material with an abrupt increase in stiffness, the 
radiation damping effects are practically negligible when the natural period of vibration of the stratum in 
shear, 
4 s 
s 
s 
T D
v 
= (C5.6-28) 
is smaller than the natural period of the flexibly supported structure, T%. The quantity Ds in this formula 
represents the depth of the stratum. It follows that the values of o presented in Figure 5.6-1 are 
applicable only when: 
s 4 s 1 
s 
T D 
T% vT% 
= = (C5.6-29) 
For 
s 4 s 1 
s 
T D 
T% vT% 
= < (C5.6-30) 
the effective value of the foundation damping factor,  o' , is less than o, and it is approximated by the 
second degree parabola defined by Eq. 5.6-10. 
For Ts /T% =1, Eq. 5.6-10 leads to o' = o whereas for Ts /T% = 0, it leads to 0' = 0, a value that clearly 
does not provide for the effects of material soil damping. It may be expected, therefore, that the 
computed values of 0' corresponding to small values of Ts /T% will be conservative. The conservatism 
involved, however, is partly compensated by the requirement that % be no less than % =  = 0.05. 
5.6.2.2 and 5.6.2.3 Vertical distribution of seismic forces and other effects. The vertical distributions 
of the equivalent lateral forces for flexibly and rigidly supported structures are generally different. 
However, the differences are inconsequential for practical purposes, and it is recommended that the same 
distribution be used in both cases, changing only the magnitude of the forces to correspond to the 
appropriate base shear. A greater degree of refinement in this step would be inconsistent with the 
approximations embodied in the requirements for rigidly supported structures. 
With the vertical distribution of the lateral forces established, the overturning moments and the torsional 
effects about a vertical axis are computed as for rigidly supported structures. The above procedure is 
applicable to planar structures and, with some extension, to three-dimensional structures. 
5.6.3 Response spectrum procedure. Studies of the dynamic response of elastically supported, 
multi-degree-of-freedom systems (Bielak, 1976; Chopra and Gutierrez, 1974; Veletsos, 1977) reveal that, 
within the ranges of parameters that are of interest in the design of structures subjected to earthquakes, 
soil-structure interaction affects substantially only the response component contributed by the 
fundamental mode of vibration of the superstructure. In this section, the interaction effects are considered 
only in evaluating the contribution of the fundamental structural mode. The contributions of the higher 
modes are computed as if the structure were fixed at the base, and the maximum value of a response 
 Structural Analysis Procedures 
97 
quantity is determined, as for rigidly supported structures, by taking the square root of the sum of the 
squares of the maximum modal contributions. 
The interaction effects associated with the response in the fundamental structural mode are determined in 
a manner analogous to that used in the equivalent lateral force procedure, except that the effective weight 
and effective height of the structure are computed so as to correspond exactly to those of the fundamental 
natural mode of the fixed-base structure. More specifically, W is computed from: 
( )2 
1 
1 2
1 
i i 
i i 
w 
W W 
w
f
f 
S 
= = 
S 
(C5.6-31) 
which is the same as Eq. 5.3-2, and h is computed from Eq. 5.6-13. The quantity fi1 in these formulas 
represents the displacement amplitude of the ith floor level when the structure is vibrating in its 
fixed-base, fundamental natural mode. The structural stiffness, k , is obtained from Eq. 5.6-4 by taking 
W=W1 and using for T the fundamental natural period of the fixed-base structure, T1. The fundamental 
natural period of the interacting system, T%1, is then computed from Eq. 5.6-3 (or Eq. 5.6-5 when 
applicable) by taking T = T1. The effective damping in the first mode, , is determined from Eq. 5.6-9 
(and Eq. 5.6-10 when applicable) in combination with the information given in Figure 5.6-1. The 
quantity h in the latter figure is computed from Eq. 5.6-13. 
With the values of 1 T% and 1 
% established, the reduction in the base shear for the first mode, .Vl, is 
computed from Eq. 5.6-2. The quantities Cs and C%s in this formula should be interpreted as the seismic 
coefficients corresponding to the periods T1 and T%1 , respectively; % should be taken equal to %1; and W 
should be determined from Eq. C5.6-31. 
The sections on lateral forces, shears, overturning moments, and displacements follow directly from what 
has already been noted in this and the preceding sections and need no elaboration. It may only be pointed 
out that the first term within the brackets on the right side of Eq. 5.6-14 represents the contribution of the 
foundation rotation. 
5.6.3.3 Design values. The design values of the modified shears, moments, deflections, and story drifts 
should be determined as for structures without interaction by taking the square root of the sum of the 
squares of the respective modal contributions. In the design of the foundation, the overturning moment at 
the foundation-soil interface determined in this manner may be reduced by 10 percent as for structures 
without interaction. 
The effects of torsion about a vertical axis should be evaluated in accordance with the requirements of 
Sec. 5.2.4 and the P-delta effects should be evaluated in accordance with the requirements of Sec. 5.2.6.2, 
using the story shears and drifts determined in Sec. 5.6.3.2. 
Other methods of considering the effects of soil-structure interaction. The procedures proposed in 
the preceding sections for incorporating the effects of soil-structure interaction provide sufficient 
flexibility and accuracy for practical applications. Only for unusual structures and only when the 
requirements indicate that the interaction effects are of definite consequence in design, would the use of 
more elaborate procedures be justified. Some of the possible refinements, listed in order of more or less 
increasing complexity, are: 
1. Improve the estimates of the static stiffnesses of the foundation, Ky and K., and of the foundation 
damping factor, 0, by considering in a more precise manner the foundation type involved, the effects 
of foundation embedment, variations of soil properties with depth, and hysteretic action in the soil. 
Solutions may be obtained in some cases with analytical or semi-analytical formulations and in others 
by application of finite difference or finite element techniques. A concise review of available 
analytical formulations is provided in Gazetas (1991). It should be noted, however, that these 
2 003 Commentary, Chapter 5 
98 
solutions involve approximations of their own that may offset, at least in part, the apparent increase in 
accuracy. 
2. Improve the estimates of the average properties of the foundation soils for the stipulated design 
ground motion. This would require both laboratory tests on undisturbed samples from the site and 
studies of wave propagation for the site. The laboratory tests are needed to establish the actual 
variations with shearing strain amplitude of the shear modulus and damping capacity of the soil, 
whereas the wave propagation studies are needed to establish realistic values for the predominant soil 
strains induced by the design ground motion. 
3. Incorporate the effects of interaction for the higher modes of vibration of the structure, either 
approximately by application of the procedures recommended in Bielak (1976), Roesset et al. (1973), 
and Tsai (1974) or by more precise analyses of the structure-soil system. The latter analyses may be 
implemented either in the time domain or by application of the impulse response functions presented 
in Veletsos and Verbic (1974). However, the frequency domain analysis is limited to systems that 
respond within the elastic range while the approach involving the use of the impulse response 
functions is limited, at present, to soil deposits that can adequately be represented as a uniform elastic 
halfspace. The effects of yielding in the structure and/or supporting medium can be considered only 
approximately in this approach by representing the supporting medium by a series of springs and 
dashpots whose properties are independent of the frequency of the motion and by integrating 
numerically the governing equations of motion (Parmelee et al., 1969). 
4. Analyze the structure-soil system by finite element method (for example, Lysmer et al., 1981; Borja 
et al., 1992), taking due account of the nonlinear effects in both the structure and the supporting 
medium. 
It should be emphasized that, while these more elaborate procedures may be appropriate in special cases 
for design verification, they involve their own approximations and do not eliminate the uncertainties that 
are inherent in the modeling of the structure-foundation-soil system and in the specification of the design 
ground motion and of the properties of the structure and soil. 
REFERENCES 
Anderson, D. G., and F. E. Richart, Jr. 1976. Effects of Straining on Shear Modulus of Clays. Journal 
of the ASCE Geotechnical Engineering Division 102 (679):975-987. 
Arias, A., and R. Husid. 1965. Influence of Damping on Earthquake Response of Structures. Revista 
del IDIEM 1 (3). In Spanish. 
Ballard, R. F., and F. G. McLean. 1975. Seismic Field Methods for In Situ Moduli. In Proceedings of 
the Conference on In Situ Measurement of Soil Properties, Vol. I, pp. 121-150. North Carolina State 
University. 
Bernal, D. 1987. Amplification Factors for Inelastic Dynamic P-delta Effects in Earthquake Analysis. 
Earthquake Engineering and Structural Dynamics, 15:835-881. 
Bielak, J. 1976. Modal Analysis for Building-Soil Interaction. Journal of the ASCE Engineering 
Mechanics Division 102 (EM5):771-786. 
Bielak, J. 1975. Dynamic Behavior of Structures with Embedded Foundations. In Earthquake 
Engineering and Structural Dynamics, Vol. 3, pp. 259-274. 
Borja, R. I., H. A. Smith, W-H. Wu, and A. P. Amies. 1992. A methodology for nonlinear soil-structure 
interaction effects using time-domain analysis techniques, Report 101. Stanford, California: Blume 
Earthquake Engineering Center. 
Chang, G.A. and Mander, J.B., Seismic Energy Based Fatigue Damage Analysis of Bridge Columns: 
Part II-Evaluation of Seismic Demand, NCEER-94-0013, National Center for Earthquake Engineering 
Research, State University of New York at Buffalo, 1994. 
 Structural Analysis Procedures 
99 
Chopra, Anil K., Dynamics of Structures: Theory and Application to Earthquake Engineering, 
Prentice-Hall, Englewood Cliffs, New Jersey, 1995. 
Chopra, A. K., and J. A. Gutierrez. 1974. Earthquake Analysis of Multistory Buildings Including 
Foundation Interaction. Journal of Earthquake Engineering and Structural Dynamics 3:65-67. 
Clough, R. W. and J. Penzien. 1975. Dynamics of Structures. New York: McGraw-Hill. 
Elsabee, F., I. Kausel, and J. M. Roesset. 1977. Dynamic Stiffness of Embedded Foundations. In 
Proceedings of the ASCE Second Annual Engineering Mechanics Division Specialty Conference, 
pp. 40-43. 
Erden, S. M. 1974. Influence of Shape and Embedment on Dynamic Foundation Response. Thesis 
submitted in partial fulfillment of the requirements for the degree of Doctor of Philosophy, University of 
Massachusetts, Amherst. 
Gazetas, G. 1991. Foundation Vibrations, in Foundation Engineering Handbook, 2nd edition, edited 
by H.-Y. Fang. 
Goel, R. K., and A. K. Chopra. 1998. Period Formulas for Concrete Shear Wall Buildings. Journal of 
Structural Engineering, ASCE, April. 
Goel, R. K., and A. K. Chopra. 1997. Period Formulas for Moment-Resisting Frame Buildings. 
Journal of Structural Engineering, ASCE, November. 
Hara, A., T. Ohta, M. Niwa, S. Tanaka, and T. Banno. 1974. Shear Modulus and Shear Strength of 
Cohesive Soils. Journal of the Japanese Society of Soil Mechanics and Foundation Engineering 14 
(3):1-24. 
Hardin, B. O. 1978. The Nature of Stress-Strain Behavior of Soils, Earthquake Engineering and Soil 
Dynamics, ASCE, Vol. I, 3-90. 
Hardin, B. O., and V. P. Drnevich. 1975. Shear Modulus and Damping in Soils: Design Equations and 
Curves. Journal of the ASCE Soil Mechanics and Foundation Division 98 (SM7):667-695. 
Jennings, P. C., and J. Bielak. 1973. Dynamics of Building-Soil Interaction. Bulletin of Seismology 
Society of America 63 (1):9-48. 
Kausel, E. 1974. Forced Vibrations of Circular Foundations on Layered Media, Report R74-11. 
Cambridge: Massachusetts Institute of Technology. 
Kausel, E., and J. M. Roesset. 1975. Dynamic Stiffness of Circular Foundations. Journal of the ASCE 
Engineering Mechanics Division 101 (EM6):771-785. 
Kramer, S. L. 1996. Geotechnical Earthquake Engineering. Upper Saddle River, New Jersey: 
Prentice-Hall. 
Kuribayashi, E., T. Iwasaki, and F. Tatsuoka. 1974. Effects of Stress Conditions on Dynamic Properties 
of Sands. Bulletin of the International Institute of Seismology and Earthquake Engineering 12:117-130. 
Liu, S. C., and L. W. Fagel. 1971. Earthquake Interaction by Fast Fourier Transform. Journal of the 
ASCE Engineering Mechanics Division 97 (EM4):1223-1237. 
Luco, J. E. 1974. Impedance Functions for a Rigid Foundation on a Layered Medium. Nuclear 
Engineering and Design (31):204-217. 
Lysmer, J., M.Tabatabaie-Raissi, F. Tajirian, S. Vahdani, and F. Ostadan. 1981. SASSI: A System for 
Analysis of Soil-Structures Interaction, Report UCB/GT-81/02. Berkeley: University of California. 
Miranda, E. and Bertero, V.V., Evaluation of Strength Reduction Factors for Earthquake Resistant 
Design, Earthquake Spectra, Vol. 10, No. 2, pp. 357-380, 1994. 
Nair, K., H. Gray, and N. C. Donovan. 1969. Analysis of Pile Group Behavior. ASTM Special 
Technical Publication 44. 
2 003 Commentary, Chapter 5 
100 
Nassar, A.A., and Krawinkler, H., Seismic Demands for SDOF and MDOF Systems, Report No. 95, the 
John A. Blume Earthquake Engineering Center, Stanford University, Stanford, California, June, 1991. 
Newmark, N. M., J. A. Blume, and K. K. Kapur. 1973. Seismic Design Spectra for Nuclear Power 
Plants. Journal of the ASCE Power Division 99 (PO 2):873-889. 
Newmark, N. M., and E. Rosenblueth. 1971. Fundamentals of Earthquake Engineering. New York: 
Prentice-Hall. 
Novak, M. 1974. Effect of Soil on Structural Response to Wind and Earthquake. In Earthquake 
Engineering and Structural Dynamics, Vol. 3, pp. 79-96. 
Novak, M., and Y. O. Beredugo. 1975. Vertical Vibrations of Embedded Footings. Journal of the 
ASCE Soil Mechanics and Foundations Division 98 (SM12):1291-1310. 
Novak, M. 1974. Dynamic Stiffness and Damping of Piles. Canadian Geotechnical Journal 
11:574-598. 
Parmelee, R. A., D. S. Perelman, and S. L. Lee. 1969. Seismic Response of Multistory Structures on 
Flexible Foundations. Bulletin of the Seismological Society of America 29:1061-1070. 
Poulos, H. G. and E. H. Davis. 1974. Elastic Solutions for Soil and Rock Mechanics. New York: Wiley 
and Sons. 
Rainer, J. H. 1975. Simplified Analysis of Dynamic Structure-Ground Interaction. Canadian Journal 
of Civil Engineering 2 (3):345-356. 
Rainer, J. H. 1975. Damping in Dynamic Structure-Foundation Interaction. Canadian Geotechnical 
Journal 12:13-25. 
Richart, F. E., J. R. Hall, Jr., and R. D. Woods. Vibrations of Soils and Foundations. Englewood Cliffs, 
New Jersey: Prentice-Hall. 
Roesset et al. 1973. 
Rutenburg, A. 1985. Simplified P-delta Analysis for Asymmetric Structures. Journal of the Structural 
Division, ASCE, 108(ST9). 
Seed, H. B., R. T. Wong, I. M. Idriss, and K. Tokimatsu. 1984. Moduli and Damping Factors for 
Dynamic Analyses of Cohesionless Soils. Rpt. No. UCB/EERC-84/14, Earthquake Engineering 
Research Center, Univ. of California, Berkeley. 
Seed, H. B., J. Lysmer, and R. Hwang. 1974. Soil-Structure Interaction Analysis for Seismic 
Response. Journal of the ASCE Geotechnical Engineering Division 101 (GT5):439-457. 
Seed, H. B., R. V. Whitman, and J. Lysmer. 1977. Soil-Structure Interaction Effects in the Design of 
Nuclear Power Plants. In Structural and Geotechnical Mechanics, A Volume Honoring N. M. Newmark, 
edited by W. J. Hall. Englewood Cliffs, New Jersey: Prentice-Hall. 
Stokoe, K. H., II, and R. D. Woods. 1975. In Situ Shear Wave Velocity by Cross-Hole Method. 
Journal of the ASCE Soil Mechanics and Foundations Division 98 (SM5):443-460. 
Structural Engineers Association of California. 1968, 1973, 1974, 1976, and 1987. Recommended 
Lateral Force Requirements and Commentary. San Francisco: SEAOC. 
Thomson, W. T. 1965. Vibration Theory and Application. New York: Prentice-Hall. 
Tomlinson, M..J. 1994. Pile Design and Construction Practice, E&FN Spon, London, UK, 4th Edition. 
Tsai, N. C. 1974. Modal Damping for Soil-Structure Interaction. Journal of the ASCE Engineering 
Mechanics Division 100 (EM2):323-341. 
Vaish, A. K., and A. K. Chopra. 1974. Earthquake Finite Element Analysis of Structure-Foundation 
Systems. Journal of the ASCE Engineering Mechanics Division 100 (EM6):1011-1016. 
 Structural Analysis Procedures 
101 
Veletsos, A. S., and V. V. Nair. 1975. Seismic Interaction of Structures on Hysteretic Foundations. 
Journal of the ASCE Structural Division 101 (ST1):109-129. 
Veletsos, A. S. 1977. Dynamics of Structure-Foundation Systems. In Structural and Geotechnical 
Mechanics, A Volume Honoring N. M. Newmark, edited by W. J. Hall, pp. 333-361. Englewood Cliffs, 
New Jersey: Prentice-Hall. 
Veletsos, A. S., and J. W. Meek. 1974. Dynamic Behavior of Building Foundation Systems. 
Earthquake Engineering and Structural Dynamics 3 (2):121-138. 
Veletsos, A. S., and B. Verbic. 1973. Vibration of Viscoelastic Foundations. Earthquake Engineering 
and Structural Dynamics 2 (1): 87-105. 
Veletsos, A. S., and Y. T. Wei. 1971. Lateral and Rocking Vibration of Footings. Journal of the 
ASCE Soil Mechanics and Foundations Division 97 (SM9):1227-1248. 
Veletsos, A. S., and B. Verbic. 1974. Basic Response Functions for Elastic Foundations. Journal of 
the ASCE Engineering Mechanics Division 100 (EM2):189-205. 
Vidic, T., Fajfar, P., and Fischinger, M. 1992. A Procedure for Determining Consistent Inelastic Design 
Spectra, Proc. Workshop of Nonlinear Seismic Analysis of RC Structures, Bled, Slovenia, July, 1992. 
Vucetic, M., and R. Dobry. 1991. Effect of Soil Plasticity on Cyclic Response. J. Geotech. Engrg. 
ASCE, 117(1), 89-107. 
Wallace, J. W., J. P. Stewart, and A. S. Whittaker, editors. 1999. Building Vulnerability Studies: 
Modeling and Evaluation of Tilt-up and Steel Reinforced Concrete Buildings, Report to PEER-PG&E 
Cooperative Research Program. Los Angeles: University of California Department of Civil and 
Environmental Engineering. 
Wiegel, R. L., Ed. 1970. Earthquake Engineering. New York: Prentice-Hall. 
Wilson, E. L., and A. Habibullah. 1987. Static and Dynamic Analysis of Multi-story Buildings 
Including P-delta Effects. Earthquake Spectra 3(2). 
2 003 Commentary, Chapter 5 
102 
This page is intentionally left blank. 
 103 
Appendix to Chapter 5 
NONLINEAR STATIC PROCEDURE 
A5.2 NONLINEAR STATIC PROCEDURE 
The nonlinear static procedure is intended to provide a simplified approach for directly determining the 
nonlinear response behavior of a structure at different levels of lateral displacements, ranging from initial 
elastic response through development of a failure mechanism and initiation of collapse. Response 
behavior is gauged by measurement of the strength of the structure, at various increments of lateral 
displacement. 
Usually the shear resisted by the system at yield of the first element of the structure is defined as the 
elastic strength, although this may not correspond to yield of the entire structure. When traditional 
linear methods of design (using R factors) are employed, this elastic strength will not be less than the 
design base shear. 
If a structure is subjected to larger lateral loads than that represented by the elastic strength, a number of 
elements will yieldeventually forming a mechanism. For most structures, multiple configurations of 
mechanisms are possible. The mechanism caused by the smallest set of forces is likely to appear before 
others do. That mechanism is considered to be the dominant mechanism. Standard methods of plastic or 
limit analysis can be used to determine the strength corresponding to such mechanisms. However, such 
limit analysis cannot determine the deformation at the onset of such a mechanism. If the yielding 
elements are able to strain harden, the mechanism will not allow an increase of deformations without 
some increase of lateral forces and the mechanism is stable. Moreover, it can be considered as a flexible 
version of the original frame structure. Figure CA5.2-1, which shows a plot of the lateral structural 
strength vs. deformation (or pushover curve) for a hypothetical structure, illustrates these concepts. 
Figure CA5.2-1 Strength-deformation relation for a frame structure 

2003 Commentary, Appendix to Chapter 5 
104 
If, after the structure develops a mechanism, it deforms an additional substantial amount, elements within 
the structure may fail (fracture, buckle, etc.) and thus cease to contribute strength to the structural system. 
In such cases, the strength of the structure will diminish with increasing deformation. In the event of 
failure of an essential element, or group of elements, the entire structure may lose capacity to carry the 
gravity or lateral loads. Such loss of load-carrying capacity can also occur if the lateral deformation 
becomes so great that the P-delta effects exceed the residual lateral stiffness of the structure. Such 
conditions are defined as collapse and the deformation associated with collapse is defined as the ultimate 
deformation. This deformation can be determined by the nonlinear static procedure and also by plastic 
or limit analysis. 
As shown in Figure CA5.2-1, many structures exhibit a range of behavior between the development of 
first yielding and development of a mechanism. When the structure deforms while elements are yielding 
sequentially (shown as progressive yielding), the relation between external forces and deformations 
cannot be determined by simple limit analysis. For such a case, other methods of analysis are required. 
The purpose of nonlinear static procedure is to provide a simplified method of determining structural 
response behavior at deformation levels between those that can be conveniently analyzed using limit state 
methods. 
A5.2.1 Modeling. In this procedure, the structure is modeled using elements having stiffness properties 
that are dependent on the amount of deformation imposed on the element. All elements than can be 
subjected to deformations or forces larger than those corresponding to yield should be modeled with 
nonlinear properties. At a minimum, nonlinear stiffness properties (using a bilinear model) should 
include initial elastic stiffness, yield strength (and yield deformation), and post-yield characteristics 
including the point of loss of strength (and associated deformation) or point of complete fracture or loss 
of stability. 
A5.2.2 Lateral loads. The analysis is performed by applying a incrementally increasing pattern of 
lateral loads distributed throughout the structure. The analysis traces the internal distribution of loads and 
deformations as the load amplitude is progressively increased. Moreover it records the strengthdeformation 
relation and the characteristic events occurring as the analysis progresses. The strengthdeformation 
relation typically takes a shape similar to that shown in Figure CA5.2-1. 
It should be noted that nonlinear static analysis can be used to determine the order of yielding of elements 
in the progressive yielding range (see Figure CA5.2-1) and the associated strengths and deformations. 
The analysis can also identify the deformations associated with fractures or failure of components and the 
entire structure. However, it is accurate only if the applied pattern of loads induces a pattern of 
deformation in the structure that is similar to that which will be induced by the earthquake ground motion. 
This can be controlled, to some extent, through application of an appropriate pattern of loads. However, 
this method is generally limited in applicability to structures that have limited higher-mode participation. 
The force-deformation sequence predicted by the analysis is a function of the configuration of the set of 
monotonically increasing loads. In order to capture the dynamic behavior of the structure, the forcedeformation 
relation should be properly defined as the instantaneous distribution of inertial forces when 
the maximum response of structure occurs. Therefore, the load configuration should be redefined at each 
point on the pushover curve, proportional to the instantaneous configuration of inertial forces. Such a 
configuration is dependent on the instantaneous modal characteristics of the structure and their 
combination. Since the structure is nonlinear, the instantaneous modal characteristics depend on the 
modified properties due to inelastic deformations, affecting the load distribution at each step, accordingly. 
Such use of a varying, deformation-dependent load configuration would require almost as much labor and 
uncertainty as application of a full nonlinear response history procedure. Such effort would be 
inappropriate for the simplified approach that the nonlinear static procedure is intended to provide. 
Therefore, the load configuration and intensity are approximated in the nonlinear static procedures. 
Several approximations are available, including the following: 
1. An approximate distribution proportional to the idealized elastic response model as used in the 
equivalent lateral force procedure: 
 Nonlinear Static Procedure 
105 
k 
i i 
i k 
j i 
j
F w h V 
w h 
= S (CA5.2-1) 
where, F, w, h and V are the story inertia force, story weight, story height, and base shear, 
respectively; k is a coefficient ranging between 1 and 2, as defined in Provisions Sec. 5.2.3. 
2. A better approximation, using the dominant mode of vibration (such as the first mode in moderate 
height building structures): 
i i 
i 
i i 
i
F w V 
w
f
f 
= S (CA5.2-2) 
where, fi 
is the dominant mode shape. This approximation allows the three-dimensional distribution 
of inertia forces to be obtained when such considerations are important. 
3. A still more complete approximation using several significant modes of vibration. In such cases the 
modes for which the total equivalent modal mass exceed 90 percent should be included. The load 
configuration is given by: 
( )( ) 
( ) ( ) 
2 1/ 2 
2 2 1/ 2 
i d ai ad 
i id 
i 
i id 
i d ai ad 
F w S S V 
w 
S S 
f
f 
= ...S ..G G ..... 
S ...S..G G ..... 
(CA5.2-3) 
where, Gi and Sai are the modal participation factor and the spectral acceleration, respectively, and 
subscript d indicates the dominant mode. ( Gi= Swifi; where the mode shapes, f, are mass 
normalizedthat is Swifi2 g=1.) 
4. An approximation that takes into account both higher mode contributions and changes in the loading 
due to yielding of the structure. In this case the load configuration described by Eq. CA5.2-3) is 
calculated and reevaluated when the modal characteristics of the structure change as it yields. Such 
procedure has also termed an adaptive push-over analysis. 
The Provisions adopt the simplest of these approaches, indicated as item 1 above, though use of the more 
complex approaches is not precluded. Nonlinear static analysis options exist in several commercially 
available and public-domain analysis platforms. 
A5.2.3 Target displacement. The nonlinear analysis should be continued by increasing the amplitude of 
the pattern of lateral loads until the deflections at the control point exceeds 150 percent of the target 
displacement. The expected inelastic deflection at each level shall be determined by combining the 
elastic modal values as obtained from Sec. 5.3.5 and 5.3.6 multiplied by the factor 
(1 s 1) ( ) 
s 1 
d 
T T 
C TT 
R 
- 
= + (CA5.2-4) 
where Ts is the characteristic period of the response spectrum, defined as the period associated with the 
transition from the constant-acceleration segment of the spectrum to the constant-velocity segment of the 
spectrum and Rd is the ratio of the total design base shear to the fully yielded strength of the major 
mechanism, which can be obtained according to Rd = R/Oo, with R and Oo given in Table 4.3-1. The 
combination shall be carried out by taking the square root of the sum of the squares of each of the modal 
values or by the complete quadratic combination technique. 
The recommendation linking the expected inelastic deformation to the elastic is based on an approach 
originally suggested by Newmark and on later studies by several other researchers. These are described 
below. 
2 003 Commentary, Appendix to Chapter 5 
106 
In a 1991 study, Nassar and Krawinkler published simplified expressions that were derived from a study 
of mean strength reduction factors computed from fifteen ground motions recorded in the Western United 
States. The records used were obtained at alluvium and rock sites. The influence of the site conditions 
was not explicitly considered. The sensitivity of mean strength reduction factors to epicentral distance, 
yield level, strain-hardening ratio, and stiffness degradation was examined. The study concluded that 
epicentral distance and stiffness degradation have negligible influence on strength reduction factors and 
proposed the following relationship for the ratio of inelastic displacements to displacements predicted by 
elastic analysis: 
11( c1)/ 1 
Rd r r 
c 
=...+ - ... = 
(CA5.2-5) 
where, 
1 
a 
a 
c T b
T T 
= + 
+ 
(CA5.2-6) 
In the above, T, is the period of vibration of the structure and r is the strength ratio. Rd is defined above. 
In 1994, Chang and Mander performed analytical studies based on an envelope of five recorded ground 
motions. The following inelastic dynamic magnification factor that relates the maximum inelastic 
displacement to the elastic spectral displacement was obtained. 
11 11 
n 
PV 
D 
R T 
r T r 
=.. - .... .. + = 
. .. . 
(CA5.2-7) 
where TPV is the period at which the maximum spectral velocity response occurs, and 
n = 1.2 + 0.025r for TPV # 1.2 sec. (CA5.2-8) 
n = 1.2 for TPV > 1.2 sec. (CA5.2-9) 
In 1992, Vidic, Fajfar, and Fischinger recommended simplified expressions derived from the study of the 
mean strength reduction factors computed from twenty ground motions recorded in the Western United 
States as well as in the 1979 Montenegro, Yugoslavia, earthquake. Systems with bilinear and stiffness 
degrading (Q-model) hysteric behavior and viscous damping proportional to the mass and the 
instantaneous stiffness were considered, resulting in the following expression: 
11 0 11 
D 
R T 
r T r 
=.. - .. + = 
. . 
(CA5.2-10) 
where T is the dominant period of structure, T0 = 0.650.3T1, and 
2 ev 
1 
ea 
T V
A 
f 
p 
f 
= (CA5.2-11) 
where V and A are the peak ground velocity and peak ground acceleration, respectively. For the 20 
ground motions considered in the study, the mean amplification factors fea and fev are 2.5 and 2.0, 
respectively. 
Miranda and Bertero (1994) suggested simplified expressions derived from the study of the mean strength 
reduction factors computed from 124 ground motions recorded on a wide range of soil conditions. The 
study considered 5-percent-damped bilinear systems undergoing displacement ductility ratios between 2 
and 6. Based on the local site conditions at the recording station, ground motions were classified into 
three groups: rock sites, and soft soil sites. In addition to the influence of soil conditions, the study 
considered the influence of magnitude and epicentral distance on strength reduction factors. The study 
concluded that soil conditions influence the reduction factors significantly (particularly for soft soil sites) 
 Nonlinear Static Procedure 
107 
and that magnitude and epicentral distance have a negligible effect on mean strength reduction factors. 
The study produced the following expression for the mean strength reduction factor: 
RD 1 1 1 
r r 
=.. - ..F+ 
. . 
(CA5.2-12) 
with, 
2 1 1 1exp3ln3 
10 2 2 5 
T 
T T T 
= + - - ....- ... - ....... 
F (CA5.2-13) 
2 1 1 2exp2ln1 
12 5 5 
T 
T T T 
F = + - - ....- ... - ....... 
(CA5.2-14) 
(CA5.2-15) 
where T is the period of vibration of the structure and Tg is the characteristic ground motion period. 
The recommended formulation contained in the Provisions is a combination of the recommendations of 
Krawinkler et al and of Vidic et al with some simplification. The Provisions require that the analysis be 
continued until the deflection at the control point exceeds 150 percent of the target displacement in order 
to account for inaccuracy due to this simplification and because small variations in strength (due to 
modeling or due to imprecise construction) can lead to large displacement variations in the inelastic 
range. 
A5.2.5 Design review. See Commentary Sec. 5.5.4. 

2 003 Commentary, Appendix to Chapter 5 
108 
This page is intentionally left blank. 
 109 
Chapter 6 Commentary 
ARCHITECTURAL, MECHANICAL, AND ELECTRICAL COMPONENT 
DESIGN REQUIREMENTS 
6.1 GENERAL 
6.1.1 Scope. The general requirements establish minimum design levels for architectural, mechanical, 
electrical, and other nonstructural systems and components recognizing occupancy use, occupant load, 
need for operational continuity, and the interrelation of structural and architectural, mechanical, 
electrical, and other nonstructural components. Several classes of components are not subject to the 
Provisions: 
1. All components in Seismic Design Category A are exempted because of the lower seismic input for 
these items. 
2. All mechanical and electrical components in Seismic Design Categories B and C are exempted if 
they have an importance factor (Ip) equal to 1.00 because of the low acceleration and the 
classification that they do not contain hazardous substances and are not required to function to 
maintain life safety. 
3. All components in all Seismic Design Categories, weighing less than 400 pounds (1780 N), and 
mounted 4 ft (1.22 m) or less above the floor are exempted if they have an importance factor (Ip) 
equal to 1.00, because they do not contain hazardous substances, are not required to function to 
maintain life safety, and are not considered to be mounted high enough to be a life-safety hazard if 
they fall. 
Storage racks are considered nonbuilding structures and are covered in Provisions Chapter 14. See 
Commentary Sec. 14.3.5. 
Storage tanks are considered nonbuilding structures and are covered in Provisions Chapter 14. See 
Commentary Sec. 14.4.7. 
When performing seismic design of nonstructural components, be aware that there may be important 
non-seismic requirements outside the scope of the building code, that may be affected by seismic 
bracing. For example, thermal expansion is often a critical design consideration in pressure piping 
systems, and bracing must be arranged in a manner that accommodates thermal movements. The design 
for seismic loads should not compromise the functionality, durability, or safety of the overall system, 
and this may require substantial collaboration and cooperation between the various disciplines in the 
design team. In some cases, such as essential facilities or hazardous environments, it may be 
appropriate to consider performance levels higher than what is required by the building code (for 
example, operability of a piping system, rather than leak tightness). 
For some components, such as exterior walls, the wind design forces may be higher than the seismic 
design forces. Even when this occurs, the seismic detailing requirements may still govern the overall 
structural design. Whenever this is a possibility, it should be investigated early in the structural design 
process. 
6.2 GENERAL DESIGN REQUIREMENTS 
6.2.2 Component importance factor. The component importance factor (Ip) represents the greater of 
the life-safety importance of the component and the hazard-exposure importance of the structure. This 
2 003 Commentary, Chapter 6 
110 
factor indirectly accounts for the functionality of the component or structure by requiring design for a 
higher force level. Use of higher Ip requirements together with application of the requirements in Sec. 
6.4.2 and 6.4.3 should provide better, more functional component. While this approach will provide a 
higher degree of confidence in the probable seismic performance of a component, it may not be 
sufficient for all components. For example, individual ceiling tiles may still fall from the ceiling grid. 
Seismic qualification approaches presently in use by the Department of Energy (DOE) and the Nuclear 
Regulatory Commission (NRC) should be considered by the registered design professional and/or the 
owner when the consequences of failure would be unacceptable. 
Components that could fall from the structure are among the most hazardous building components in an 
earthquake. These components may not be integral with the structural system and may cantilever 
horizontally or vertically from their supports. Critical issues affecting these components include their 
weight, their attachment to the structure, their breakage characteristics (glass) and their location (over an 
entry or exit, public walkway, atrium, or lower adjacent structure). Examples of items that may pose a 
falling hazard include parapets, cornices, canopies, marquees, glass, and precast concrete cladding 
panels. In addition, mechanical and electrical components may pose a falling hazard (for example, a 
rooftop tank or cooling tower, which if separated from the structure would fall to the ground). 
Special consideration should be given to components that could block means of egress or exitways if 
they were to fall during an earthquake. The term means of egress has been defined in the same way 
throughout the country, since egress requirements have been included in building codes because of fire 
hazard. The requirements for exitways include intervening aisles, doors, doorways, gates, corridors, 
exterior exit balconies, ramps, stairways, pressurized enclosures, horizontal exits, exit passageways, exit 
courts, and yards. Example items that should be included when considering egress include walls around 
stairs and corridors, and veneers, cornices, canopies, and other ornaments above building exits. In 
addition, heavy partition systems vulnerable to failure by collapse, ceilings, soffits, light fixtures, or 
other objects that could fall or obstruct a required exit door or component (rescue window or fire 
escape) could be considered major obstructions. Examples of components that do not pose a significant 
falling hazard include fabric awnings and canopies and architectural, mechanical, and electrical 
components which, if separated from the structure, will fall in areas that are not accessible (in an atrium 
or light well not accessible to the public, for instance). 
In Sec. 1.2.1 the intent is that Group III structures shall, in so far as practical, be provided with the 
capacity to function after an earthquake. To facilitate this, all nonstructural components and equipment 
in structures in Seismic Use Group III, and in Seismic Design Category C or higher, should be designed 
with an Ip equal to 1.5. All components and equipment are included because damage to vulnerable 
unbraced systems or equipment may disrupt operations following an earthquake, even if they are not 
life-safety items. Nonessential items can be considered black boxes. There is no need for 
component analysis as discussed in Sec. 6.4.2 and 6.4.3, since operation of these secondary items is not 
critical to the post-earthquake operability of the structure. Instead, the design may focus on their 
supports and attachments. 
6.2.3 Consequential damage. Although the components included in Tables 6.3-1 and 6.4-1 are listed 
separately, significant interrelationships exist among them and should not be overlooked. For example, 
exterior, nonstructural, spandrel walls may shatter and fall on the streets or walks below, seriously 
hampering accessibility and egress functions. Further, the rupture of one component could lead to the 
failure of another that is dependent on the first. Accordingly, the collapse of a single component 
ultimately may lead to the failure of an entire system. Widespread collapse of suspended ceilings and 
light fixtures in a building may render an important space or major exit stairway unusable. 
Consideration also was given to the design requirements for these components to determine how well 
they are conceived for their intended functions. Potential beneficial and/or detrimental interactions with 
the structure were examined. The interrelationship between components and their attachments were 
surveyed. Attention was given to the performance relative to each other of architectural, mechanical, 
Architectural, Mechanical, and Electrical Component Design Requirements 
111 
and electrical components; building products and finish materials; and systems within and without the 
building structure. It should be noted that the modification of one component in Table 6.3-1 or 6.4-1 
could affect another and, in some cases, such a modification could help reduce the risk associated with 
the interrelated unit. For example, landscaping barriers around the exterior of certain buildings could 
decrease the risk due to falling debris although this should not be interpreted to mean that all buildings 
must have such barriers. 
The design of components that are in contact with or in close proximity to structural or other 
nonstructural components must be given special study to avoid damage or failure when seismic motion 
occurs. An example is where an important element, such as a motor generator unit for a hospital, is 
adjacent to a non-load-bearing partition. The failure of the partition might jeopardize the motor 
generator unit and, therefore, the wall should be designed for a performance level sufficient to ensure its 
stability. 
Where nonstructural wall components may affect or stiffen the structural system because of their close 
proximity, care must be exercised in selecting the wall materials and in designing the intersection details 
to ensure the desired performance of each component. 
6.2.4 Flexibility. In the design and evaluation of support structures and the attachment of architectural 
components, flexibility should be considered. Components that are subjected to seismic relative 
displacements (that is, components that are connected to both the floor and ceiling level above) should 
be designed with adequate flexibility to accommodate imposed displacements. This is covered in Sec. 
6.2.7. In the design and evaluation of equipment support structures and attachments, flexibility will 
reduce the fundamental frequency of the supported equipment and increase the amplitude of its induced 
relative motion. This lowering of the fundamental frequency of the supported component often will 
bring it into the range of the fundamental frequency of the supporting building or into the high energy 
range of the input motion. In evaluating the flexibility/stiffness of the component attachment, the 
effects of flexibility in the load path of the components should be considered especially in the region 
near the anchor points. 
6.2.5 Component force transfer. It is required that components be attached to the structure and that 
all the required attachments be fully detailed in the design documents, or be specified in accordance 
with approved standards. These details should take into account the force levels and anticipated 
deformations expected or designed into the structure. For the purposes of the load path check, it is 
essential that detailed information concerning the components, including size, weight, and location of 
component anchors, be communicated to the registered design professional responsible for the structure 
during the design process. 
The calculation of forces as prescribed in Sec. 6.2.6 recognizes the unique dynamic and structural 
characteristics of the components as compared to structures. Components typically lack the desirable 
attributes of structures (such as ductility, toughness, and redundancy) that permit the use of greatly 
reduced lateral design forces. This is reflected in the lower values for Rp given in Tables 6.3-1 and 6.4- 
1, as compared to R values for structures. In addition, components may exhibit unique dynamic 
amplification characteristics, as reflected in the values for ap in Tables 6.3-1 and 6.4-1. Thus, for the 
calculation of the component integrity and connection to the supporting structure, greater forces are 
used, as a percentage of component mass, than are typically calculated for the overall seismic-forceresisting 
system. It is the intent of this provision that component forces be accommodated in the design 
of the structure as required to prevent local overstress of the immediate vertical and lateral load-carrying 
systems. Inasmuch as the component masses are included, explicitly or otherwise, in the design of the 
seismic-force-resisting system, it is generally sufficient for verification of a complete load path to check 
only for local overstress conditions in the vicinity of the component in question. One approach to 
achieve this is to check the capacity of the first structural element in the load path (for example, the 
floor beam directly under a component) for combined dead, live, operating, and seismic loads, using the 
horizontal and vertical loads from Sec. 6.2.6 for the seismic demand. This procedure is repeated for 
2 003 Commentary, Chapter 6 
112 
each structural element or connection in the load path until the load case including horizontal and 
vertical loads from Sec. 6.2.6 no longer governs the design of the element. This will occur when the 
component design loads generated by Sec. 6.2.6 become small relative to the dead and live load 
demands on the structural element. Where component forces have increased due to the nature of the 
anchorage system, these load increases, which take the form of reductions in Rp, or increase s in Fp, need 
not be considered in the check of the load path. 
An area of concern that is often overlooked is the reinforcement and positive connection of 
housekeeping slabs to the supporting structure. Lack of such reinforcement and connections has led to 
costly failures in past earthquakes. Therefore, the housekeeping slabs must be considered as part of the 
continuous load path be adequately reinforced, and be positively fastened to the supporting structure. 
The exact size and location of loads might not be known until the component is ordered. Therefore, the 
designer should make conservative assumptions in the design of the supporting structural elements. The 
design of the supporting structural elements must be checked once the final magnitude and location of 
the design loads have been established. 
If an architectural component were to fail during an earthquake, the mode of failure probably would be 
related to faulty design of the component, interrelationship with another component that fails, 
interaction with the structural framing, deficiencies in its type of mounting, or inadequacy of its 
attachments or anchorage. The last is perhaps the most critical when considering seismic safety. 
Building components designed without any intended structural functionsuch as infill wallsmay 
interact with the structural framing and be forced to act structurally as a result of excessive building 
deformation. The build up of stress at the connecting surfaces or joints may exceed the limits of the 
materials. Spatial tolerances between such components thus become a governing factor. These 
requirements therefore emphasize the ductility and strength of the attachments for exterior wall elements 
and the interrelationship of elements. 
Traditionally, mechanical equipment that does not include rotating or reciprocating components (such as 
tanks and heat exchangers) is anchored directly to the building structure. Mechanical and electrical 
equipment containing rotating or reciprocating components often is isolated from the structure by 
vibration isolators (such as rubber-in-shear, springs, or air cushions). Heavy mechanical equipment 
(such as large boilers) often is not restrained at all, and electrical equipment other than generators, 
which are normally isolated to dampen vibrations, usually is rigidly anchored (for example, switchgear 
and motor control centers). The installation of unattached mechanical and electrical equipment should 
be virtually eliminated for buildings covered by the Provisions. 
Friction produced solely by the effects of gravity cannot be counted on to resist seismic forces as 
equipment and fixtures often tend to walk due to rocking when subjected to earthquake motions. This 
often is accentuated by vertical ground motions. Because such frictional resistance cannot be relied 
upon, positive restraint must be provided for each component. 
6.2.6 Seismic forces. The design seismic force is dependent upon the weight of the system or 
component, the component amplification factor, the component acceleration at point of attachment to 
the structure, the component importance factor, and the component response modification factor. 
The seismic design force equations presented originated with a study and workshop sponsored by the 
National Center for Earthquake Engineering Research (NCEER) with funding from the National 
Science Foundation (NSF) (Bachman et al., 1993). The participants examined recorded acceleration 
data in response to strong earthquake motions. The objective was to develop a supportable design 
force equation that considered actual earthquake data as well as component location in the structure, 
component anchorage ductility, component importance, component safety hazard upon separation from 
the structure, structural response, site conditions, and seismic zone. Additional studies have further 
revised the equation to its present form (Drake and Bachman, 1994 and 1995). In addition, the term Ca 
has been replaced by the quantity 0.4SDS to conform to changes in Chapter 3. BSSC Technical 
Architectural, Mechanical, and Electrical Component Design Requirements 
113 
Subcommittee 8 believes that Eq. 6.2-1, 6.2-3, and 6.2-4 achieve the objectives without unduly 
burdening the practitioner with complicated formulations. 
The component amplification factor (ap) represents the dynamic amplification of the component relative 
to the fundamental period of the structure (T). It is recognized that at the time the components are 
designed or selected, the structural fundamental period is not always defined or readily available. It is 
also recognized that the component fundamental period (Tp) is usually only accurately obtained by 
expensive shake-table or pull-back tests. A listing is provided of ap values based on the expectation that 
the component will usually behave in either a rigid or flexible manner. In general, if the fundamental 
period of the component is less than 0.06 sec, no dynamic amplification is expected. It is not the 
intention of the Provisions to preclude more accurate determination of the component amplification 
factor when reasonably accurate values of both the structural and component fundamental periods are 
available. Figure C6.2-1 is from the NCEER work and is an acceptable formulation for ap as a function 
of Tp/T. Minor adjustments in the tabulated ap values were made in the 1997 Edition to be consistent 
with the 1997 Uniform Building Code. 
The component response modification factor (Rp) represents the energy absorption capability of the 
components structure and attachments. Conceptually, the Rp value considers both the overstrength and 
deformability of the components structure and attachments. In the absence of current research, it is 
believed these separate considerations can be adequately combined into a single factor. The engineering 
community is encouraged to address the issue and conduct research into the component response 
modification factor that will advance the state of the art. These values are judgmentally determined 
utilizing the collective wisdom and experience of the responsible committee. In general, the following 
benchmark values were used: 
Rp = 1.5, low deformability element 
Rp = 2.5, limited deformability element 
Rp = 3.5, high deformability element 
Minor adjustments in the tabulated Rp values were made in the 1997 Edition to correlate with Fp values 
determined in accordance with the 1997 Uniform Building Code. Researchers have proposed a 
procedure for validating values for Rp with respect to documented earthquake performance (Bachman 
and Drake, 1996). 
2 003 Commentary, Chapter 6 
114 
Tp/T 
0.5 
1.0 
0.6 0.7 1.0 
2.0 
p 
2.5
a 
1.4 2.0 
NCEER STUDY 
1991 NEHRP 
(Reference) 
Figure C6.2-1 NCEER formulation for app as function of structural and component 
periods 
Eq. 6.2-1 represents a trapezoidal distribution of floor accelerations within the structure, linearly 
varying from the acceleration at the ground (0.4SDS) to the acceleration at the roof (1.2SDS). The ground 
acceleration (0.4SDS) is intended to be the same acceleration used as design input for the structure itself 
and includes site effects. 
Examination of recorded in-structure acceleration data in response to large California earthquakes 
reveals that a reasonable maximum value for the roof acceleration is four times the input ground 
acceleration to the structure. Earlier work (Drake and Bachman, 1996, 1995 and 1996) indicated that 
the maximum amplification factor of four seems suitable (Figure C6.2-2). However, a close 
examination of recently recorded strong motion data at sites with peak ground accelerations in excess of 
0.1g indicates that an amplification factor of three is more appropriate (Figure C6.2-3). In the lower 
portions of the structure (the lowest 20 percent of the structure), both the amplification factors of three 
and four do not bound the mean plus one standard deviation accelerations. However, the minimum 
design force in Eq. 6.2-4 provides a lower bound in this region. 
Architectural, Mechanical, and Electrical Component Design Requirements 
115 
Figure C6.2-2 Revised NEHRP equation vs (Mean + 1s) 
acceleration records -all sites 
Figure C6.2-3 Revised NEHRP equation vs (Mean +1s) 
acceleration records - sites with Agg $0.1g 
At periods greater than Ts (where Ts = SD1/SDS), the acceleration response of structures ground reduces 
because the design ground motion acceleration response spectra beyond Ts starts to reduce by the ratio 
of Ts / T (where T is the fundamental period of the primary structure). Since this reduction in the design 
forces for the primary structure is accounted for in design force equations, it is justifiable to make a 
similar type of reduction in the design forces of non-structural components. However, in observing the 
actual in-structure response spectra of acceleration recordings measured at the roof levels of buildings 
with a range of fundamental periods, this reduction in response typically begins at periods about 25 
percent greater than TS. Therefore, the transition period, Tflx at the top of the primary structure at which 
forces begin reducing by a ratio of Tp/T has been lengthened by 25 percent to account for this 
observation. At the ground level (z = 0) the adjustment is 0 percent since the effect of the structure 
response has no influence on the non-structural component response. A linear interpolation is used 
between the top and bottom of the structure. 
A lower limit for Fp is set to assure a minimal seismic design force. The minimum value for Fp was 
determined by setting the quantity apAp/Rp equal 0.3SDS for consistency with current practice. 
2 003 Commentary, Chapter 6 
116 
To meet the need for a simpler formulation, a conservative maximum value for Fp also was set. Eq. 6.2- 
3 is the maximum value for Fp determined by setting the quantity apAp/Rp equal to 4.0. Eq. 6.2-3 also 
serves as a reasonable cutoff equation to assure that the multiplication of the individual factors does 
not yield an unreasonably high design force. 
To clarify the application of vertical seismic design forces in combination with horizontal design forces 
and service loads, a cross-reference was provided to Sec. 4.2.2. The value for Fp calculated in 
accordance with Chapter 6 should be substituted for the value of QE in Sec. 4.2.2. 
For elements with points of attachment at varying heights, it is recommended that Fp be determined 
individually at each height (including minima) and the values averaged. 
Alternatively for each point of attachment a force Fp shall be determined based on Eq. 6.2-1. Minima 
and maxima in Sec. 6.2.6 must be utilized in determining each Fp. The weight Wp used in determining 
each Fp should be based on the tributary weight of the component associated with the point of 
attachment. For designing the component, the attachment force Fp should be distributed relative to the 
components mass distribution over the area used to establish the tributary weight (in the instance of tiltup 
walls, a uniform horizontal load would be applied half-way up the wall equal to Fp min.). With the 
exception of out-of-plane wall anchorage to flexible diaphragms, which is covered by Eq. 4.6-1, each 
anchorage force should be based on simple statistics determined using all the distributed loads applied 
to the complete component. Cantilever parapets that are part of a continuous element should be 
separately checked for parapet forces. 
The seismic force on any component must be applied at the center of gravity of the component and must 
be assumed to act in any horizontal direction. Vertical forces on nonstructural components are specified 
in Sec. 6.2.6. 
6.2.7 Seismic relative displacements. The seismic relative displacement equations were developed as 
part of the NCEER/NSF study and workshop described above. It was recognized that displacement 
equations were needed for use in the design of cladding, stairwells, windows, piping systems, sprinkler 
components, and other components that are connected to the structure(s) at multiple levels or points of 
connection. 
Two equations are given for each situation. Eq. 6.2-5 and Eq. 6.2-7 produce real structural 
displacements as determined by elastic analysis, with no structural response modification factor (R) 
included. Recognizing that elastic displacements are not always defined or available at the time the 
component is designed or procured, default Eq. 6.2-6 and Eq. 6.2-8, which allow the use of structure 
drift limitations, also are provided. Use of these default equations must balance the need for a timely 
component design/procurement with the possible conservatism of their use. It is the intention that the 
lesser of the paired equations be acceptable for use. 
The designer also should consider other situations where seismic relative displacements could impose 
unacceptable stresses on a component or system. One such example would be a component connecting 
two pieces of equipment mounted in the same building at the same elevation, where each piece of 
equipment has its own displacements relative to the mounting location. In this case, the designer must 
accommodate the total of the separate seismic displacements relative to the equipment mounting 
location. The height over which Dp, the displacement demand, must be accommodated is often less than 
the story height hsx, and should be carefully considered. For example, a glazing system sandwiched 
between two rigid precast concrete spandrel panels may need to accommodate the entire displacement 
demand in less than 1/3 of the story height. Similar demands can occur when pipes, ducts and conduit 
that are connected to the top of a tall component are braced to the floor or roof above. 
For some items, such as ductile piping, relative seismic displacements between support points generally 
are of more significance than forces. Piping made of ductile materials such as steel or copper can 
accommodate relative displacements by local yielding but with strain accumulations well below failure 
levels. However, components made of less ductile materials can only accommodate relative 
Architectural, Mechanical, and Electrical Component Design Requirements 
117 
displacement effects by use of flexible connections, avoiding local yielding. Further, it is the intent of 
the Provisions to consider the effects of seismic support relative displacements and displacements 
caused by seismic forces on mechanical and electrical component assemblies such as piping systems, 
cable and conduit systems, and other linear systems, and the equipment to which they attach. Impact of 
components should also be avoided although ductile materials have been shown to be capable of 
accommodating fairly significant impact loads. With protective coverings, ductile mechanical and 
electrical components and many more fragile components can be expected to survive all but the most 
severe impact loads. 
6.2.8 Component anchorage. Depending on the specifics of the design condition, ductile design of 
anchors in concrete or masonry may be intended to satisfy one or all of the following objectives: (1) to 
ensure adequate load redistribution between anchors in a group, (2) to allow for anchor overload 
without precipitous failure, and/or (3) to dissipate seismic energy. Unless specific attention is paid to the 
conditions necessary to ensure the desired hysteretic response (adequate gauge length, anchor spacing, 
edge distance, steel properties, etc.), it is not recommended that anchors be relied upon for energy 
dissipation. Inasmuch as the anchor provides the transfer of load from a relatively deformable material 
(such as steel) to a low deformability material (such as concrete or masonry), achieving deformable, 
energy-absorbing behavior in the anchor itself is often difficult. On the other hand, the concept of 
providing a fuse, or deformable link, in the load path to the anchor is encouraged. This approach allows 
the designer to provide the necessary level of ductility and overstrength in the connection while at the 
same time protecting the anchor from overload and eliminates the need to balance steel strength and 
deformability in the anchor with variable edge distances and anchor spacings. 
Previous restrictions on the anchor l/d ratio as a means of defining ductile vs. non-ductile anchors have 
been deleted from the Provisions in recognition of the difficulty in defining the conditions necessary for 
real ductile behavior. For example, a single anchor with the necessary embedment to force ductile 
failure of the anchor bolt in tension may still experience concrete fracture (a non-ductile failure mode) if 
the edge distance is small, if the anchor is placed in a group of tension-loaded anchors with reduced 
spacing, or if the anchor is loaded in shear instead of tension. In fact, many if not most anchor 
applications, such as building cladding attachments and large equipment anchorages, are subject 
primarily to shear loading. In these cases, even if the anchor steel is ductile, shear failure of the bolt may 
be non-ductile, particularly if the deformation of the anchor is constrained by rigid elements on either 
side of the joint. It is therefore left to the designer to establish the necessary criteria for ductile anchor 
failure. 
Post-installed expansion and undercut anchors may now be qualified as suitable for seismic applications 
using the testing procedures outlined in ACI 355.2-01, Evaluating the Performance of Post-Installed 
Mechanical Anchors in Concrete (355.2-01) and Commentary (355.2R-01). The design of qualified anchors 
in concrete is addressed in Sec. 9.6 of the Provisions. No such standard exists as yet for chemical anchors, 
and caution should be exercised in their use in earthquake environments, particularly with respect to the 
effects of earthquake-induced cracking of the concrete or masonry on anchor capacity. The capacity of 
anchors in masonry is rarely governed by steel capacity, and as such masonry anchors should in general be 
considered to be non-ductile. For this reason, the design of anchors in masonry should be carried out with an 
Rp of 1.5. 
For purposes of the Provisions, a chemical anchor is a post-installed anchor rod, usually steel, which is 
inserted into a drilled hole in concrete or masonry together with a polymer or cementitious grout and 
which derives its tension capacity primarily from bond. On the other hand, reference to adhesives is 
intended to include steel plates and other structural elements adhered to the surface of another structural 
component with adhesive. An example of this type of application is the attachment of computer access 
floors base plates to a floor slab with epoxy. This type of connection is typically non-ductile. 
2 003 Commentary, Chapter 6 
118 
Allowable loads for anchors should not be increased for earthquake loading. Possible reductions in 
allowable loads for particular anchor types to account for loss of stiffness and strength should be 
determined by means of appropriate dynamic testing. 
Anchors that are used to support towers, masts, and equipment often are provided with double nuts to 
allow for leveling during installation. Where baseplate grout is provided at such double-nutted anchors, 
it should not be relied upon to carry loads since it can shrink and crack or is often omitted altogether. In 
this case, the anchors are loaded in tension, compression, shear, and flexure and should be designed 
accordingly. Prying forces on anchors, which result from a lack of rotational stiffness in the connected 
part, can be critical for anchor design and must be considered explicitly. 
For anchorages that are not provided with a mechanism to transfer compression loads, the design for 
overturning must reflect the actual stiffness of the baseplate, equipment, housing, etc., in determining 
the location of the compression centroid and the distribution of uplift loads to the anchors. 
Possible reductions in allowable loads for particular anchor types to account for loss of stiffness and 
strength should be determined through appropriate dynamic testing. 
While the requirements do not prohibit the use of single anchor connections, it is considered good 
practice to use at least two anchors in any load-carrying connection whose failure might lead to 
collapse, partial collapse, or disruption of a critical inertial load path. 
Tests have shown that there are consistent shear ductility variations between bolts anchored to drilled or 
punched plates with nuts and connections using welded, shear studs. Recommendations for design are 
not presently available but this issue should be considered in critical connections subject to dynamic or 
seismic loading. 
It is important to relate the anchorage demands defined by Chapter 6 with the material capacities 
defined in the other chapters (e.g., Chapters 9 and 11). 
6.2.8.5 Power-actuated fasteners. Generally, power-actuated fasteners in concrete tend to exhibit 
variations in load capacity that are somewhat larger than post-installed drilled anchors. Therefore, the 
suitability of power-actuated fasteners should be demonstrated by a simulated seismic test program 
prior to their use. When properly installed in steel, such fasteners are reliable, showing high capacities 
with very low variability. 
6.2.9 Construction documents. The committee believes that each quality assurance activity specified 
in Chapter 2 should have a clearly defined basis. As a result, construction documents are required for 
all components for which Chapter 2 requires special inspection or testing. 
The committee believes that, in order to provide a reasonable level of assurance that the construction 
and installation of components is consistent with the basis of the supporting seismic design, appropriate 
construction documents are needed. Of particular concern are systems involving multiple trades and 
suppliers. In these cases, it is important that a registered design professional prepare construction 
documents for use by the various trades and suppliers in the course of construction. 
6.3 ARCHITECTURAL COMPONENTS 
The requirements of Sec. 6.3 are intended to reduce the threat of life safety hazards posed by 
components and elements from the standpoint of stability and integrity. There are several circumstances 
where such components may pose a threat. 
1. Where loss of integrity and/or connection failure under seismic motion poses a direct hazard in that 
the components may fall on building occupants. 
2. Where loss of integrity and/or connection failure may result in a hazard for people outside of a 
building because components such as exterior cladding and glazing may fall on them. 
3. Where failure or upset of interior components may impede access to a required exit. 
Architectural, Mechanical, and Electrical Component Design Requirements 
119 
The requirements are intended to apply to all of the circumstances listed above. Although the safety 
hazard posed by exterior cladding is obvious, judgment may be needed in assessing the extent to which 
the requirements should be applied to other hazards. 
Property loss through damage to architectural components is not specifically addressed in the 
Provisions. Function and operation of a building also may be affected by damage to architectural 
components if it is necessary to cease operations while repairs are undertaken. In general, requirements 
to improve life-safety also will reduce property loss and loss of building function. 
In general, functional loss is more likely to be affected by loss of mechanical or electrical components. 
Architectural damage, unless very severe, usually can be accommodated on a temporary basis. Very 
severe architectural damage results from excessive structural response that often also results in 
significant structural damage and building evacuation. 
6.3.1 Forces and displacements. Components that could be damaged by or could damage other 
components and are fastened at multiple locations to a structure should be designed to accommodate 
seismic relative displacements. Such components include glazing, partitions, stairs, and veneers. 
Certain types of veneer elements, such as aluminum or vinyl siding and trim, possess high 
deformability. These systems are generally light and can undergo large deformations without separating 
from the structure. However, care must be taken when designing these elements to ensure that the low 
deformability components that may be part of the curtain wall system, such as glazing panels, have been 
detailed to accommodate the expected deformations without failure. 
Specific requirements for cladding are provided. Glazing, both exterior and interior, and partitions must 
be capable of accommodating story drift without causing a life-safety hazard. Design judgment must be 
used with respect to the assessment of life-safety hazard and the likelihood of life-threatening damage. 
Special detailing to accommodate drift for typical replaceable gypsum board or demountable partitions 
is not likely to be cost-effective, and damage to these components poses a low hazard to life safety. 
Nonstructural fire-resistant enclosures and fire-rated partitions may require some special detailing to 
ensure that they retain their integrity. Special detailing should provide isolation from the adjacent or 
enclosing structure for deformation equivalent to the calculated drift (relative displacement). In-plane 
differential movement between structure and wall is permitted. Provision also must be made for out-ofplane 
restraint. These requirements are particularly important in relation to the larger drifts experienced 
in steel or concrete moment frame structures. The problem is less likely to be encountered in stiff 
structures, such as those with shear walls. 
Differential vertical movement between horizontal cantilevers in adjacent stories (such as cantilevered 
floor slabs) has occurred in past earthquakes. The possibility of such effects should be considered in the 
design of exterior walls. 
6.3.2 Exterior nonstructural wall elements and connections. The Provisions requires that 
nonbearing wall panels that are attached to or enclose the structure be designed to resist the (inertial) 
forces and to accommodate movements of the structure resulting from lateral forces or temperature 
change. The force requirements often overshadow the importance of allowing thermal movement and 
may therefore require special detailing in order to prevent moisture penetration and allow thermal 
movements. 
Connections should be designed so as to prevent the loss of load-carrying capacity in the event of 
significant yielding. Between points of connection, panels should be separated from the structure 
sufficiently to avoid contact due to seismic action. 
The Provisions requires allowance for story drift. This required allowance can amount to 2 in. (50 mm) 
or more from one floor to the next and may present a greater challenge to the designer than requirements 
for the forces. In practice, separations between adjacent panels, intended to limit contact and resulting 
panel mis-alignment and/or damage under all but extreme building response, are limited to about 3/4 in. 
2 003 Commentary, Chapter 6 
120 
(19 mm) for practical joint detailing with acceptable appearance. The Provisions calls for a minimum 
separation of 1/2 in. (13 mm). The design should be consistent with the manufacturing and construction 
tolerances of the materials used to achieve this dimension. 
If wind loads govern, connectors and panels should allow for not less than two times the story drift 
caused by wind loads determined using a return period appropriate to the site location. 
The Provisions requirements are in anticipation of frame yielding to absorb energy. Appropriate 
isolation can be achieved by means of slots, but the use of long rods that flex is preferable because this 
approach does not depend on precise installation to achieve the desired action. The rods must be 
designed to carry tension and compression in addition to induced flexural stresses. Care must be used in 
allowing inelastic bending in the rods. Threaded rods pushed into the strain-hardening region of the 
stress-strain curve are subject to brittle low-cycle fatigue failures. Floor-to-floor wall panels are usually 
rigidly attached to and move with the floor structure nearest the panel bottom. In this condition, 
isolation connections are used at the upper attachments so that panels translate with the load supporting 
structure below and are not subjected to large in-plane forces due to movement of the building. Panels 
also can be supported at the top with isolation connections at the bottom. 
When determining the length of slot or displacement demand for the connection, the cumulative effect 
of tolerances in the supporting frame and cladding panel must be considered. 
To minimize the effects of thermal movements and shrinkage on architectural cladding panels, the 
connection system is generally detailed to be statically determinate. As a result, cladding panel support 
systems often lack redundancy and failure of a single connection can have catastrophic consequences. 
In recognition of this, the Provisions require that fasteners be designed for approximately 4 times the 
required panel force and that the connecting member be ductile. This is intended to ensure that the 
energy absorption takes place in the connecting member and not at the connection itself and that the 
more brittle fasteners remain essentially elastic under seismic loading. The factor of 4 has been 
incorporated into the ap and Rp factors in consideration of installation and material variability and the 
consequences of a brittle connection failure in a statically determinate system. 
6.3.3 Out-of-plane bending. Most walls are subject to out-of-plane forces when a building is shaken 
by an earthquake. These forces and the bending they induce must be considered in the design of wall 
panels, nonstructural walls, and partitions. This is particularly important for systems composed of 
brittle materials or materials with low flexural strength. The conventional limits based upon deflections 
as a proportion of the span may be used with the applied force as derived in Sec. 6.2.6. 
Judgment must be used in assessing the deflection capability of the component. The intent is that a 
heavy material (such as concrete block) or an applied finish (such as brittle heavy stone or tile) should 
not fail in a hazardous manner as a result of out-of-plane forces. Deflection in itself is not a hazard. A 
steel-stud partition might undergo considerable deflection without creating a hazard; but if the same 
partition supports a marble facing, a hazard might exist and special detailing may be necessary. 
6.3.4 Suspended ceilings. Suspended ceiling systems usually are fabricated using a wide range of 
building materials with individual components having different material characteristics. Some systems 
are homogeneous whereas others incorporate suspension systems with acoustic tile or lay-in panels. 
Seismic performance during recent, large earthquakes in California has raised two concerns: 
1. Support of the individual panels at walls and expansion joints, and 
2. Interaction with fire sprinkler systems. 
In an attempt to address these concerns, alternate methods were developed in a cooperative effort by 
representatives of the ceiling and fire sprinkler industries and registered design professionals. It is 
hoped that future research and investigation will result in further improvements in the Provisions. 
Architectural, Mechanical, and Electrical Component Design Requirements 
121 
Consideration must be given to the placement of seismic bracing, the relation of light fixtures and other 
loads placed into the ceiling diaphragm, and the independent bracing of partitions in order to effectively 
maintain the performance characteristics of the ceiling system. The ceiling system may require bracing 
and allowance for the interaction of components. 
Dynamic testing of suspended ceiling systems constructed according to the requirements of current 
industry seismic standards (UBC Standard 25-2) performed by ANCO Engineers, Inc. (1983) has 
demonstrated that the splayed wires, even with the vertical compression struts, may not adequately limit 
lateral motion of the ceiling system due to the flexibility introduced by the straightening of the wire end 
loops. In addition, splay wires usually are installed slack to prevent unleveling of the ceiling grid and to 
avoid above-ceiling utilities. Not infrequently, bracing wires are omitted because of obstructions. 
Testing also has shown that system performance without splayed wires or struts was good if sufficient 
clearance is provided at penetrations and closure angles are wide enough. 
The lateral seismic restraint for a non-rigidly braced suspended ceiling is primarily provided by the 
ceiling coming into contact with the perimeter wall. The wall provides a large contact surface to 
restrain the ceiling. The key to good seismic performance is that the width of the closure angle around 
the perimeter is adequate to accommodate ceiling motion and that penetrations, such as columns and 
piping, have adequate clearance to avoid concentrating restraining loads on the ceiling system. The 
behavior of an unbraced ceiling system is similar to that of a pendulum; therefore, the lateral 
displacement is approximately proportional to the level of velocity-controlled ground motion and the 
square root of the suspension length. Therefore, a new section has been added that permits exemption 
from force calculations if certain displacement criteria are met. The default displacement limit has been 
determined based on anticipated damping and energy absorption of the suspended ceiling system 
assuming minimal significant impact with the perimeter wall. 
6.3.5 Access floors. Performance of computer access floors during past earthquakes and during cyclic 
load tests indicate that typical raised access floor systems may behave in a brittle manner and may 
exhibit little reserve capacity beyond initial yielding or failure of critical connections. Recent testing 
indicates that individual panels may pop out of the supporting grid during seismic motions. 
Consideration should be given to mechanically fastening the individual panels to the supporting 
pedestals or stringers in egress pathways. 
For systems with floor stringers, it is acceptable practice to calculate the seismic force, Fp, for the entire 
access floor system within a partitioned space and then distribute the total force to the individual braces 
or pedestals. Stringerless systems need to be evaluated very carefully to ensure a viable seismic load 
path. 
Overturning effects for the design of individual pedestals is a concern. Each pedestal usually is 
specified to carry an ultimate design vertical load greatly in excess of the Wp used in determining the 
seismic force Fp. It is non-conservative to use the design vertical load simultaneously with the design 
seismic force when considering anchor bolts, pedestal bending, and pedestal welds to base plates. The 
maximum concurrent vertical load when considering overturning effects is therefore limited to the value 
of Wp used in determining Fp. Slip on heads are not mechanically fastened to the pedestal shaft and 
provide doubtful capacity to transfer overturning moments from the floor panels or stringers to the 
pedestal. 
To preclude brittle failure, each element in the seismic load path must demonstrate the capacity for 
elastic or inelastic energy absorption. Buckling failure modes also must be prevented. Lesser seismic 
force requirements are deemed appropriate for access floors designed to preclude brittle and buckling 
failure modes. 
6.3.6 Partitions. Partitions are sometimes designed to run only from floor to a suspended ceiling 
which provides doubtful lateral support. Partitions subject to these requirements must have independent 
2 003 Commentary, Chapter 6 
122 
lateral support bracing from the top of the partition to the building structure or to a substructure attached 
to the building structure. 
6.3.7 Glass in glazed curtain walls, glazed storefronts, and glazed partitions. Glass performance in 
earthquakes can fall into one of four categories: 
1. The glass remains unbroken in its frame or anchorage. 
2. The glass cracks but remains in its frame or anchorage while continuing to provide a weather 
barrier, and to be otherwise serviceable. 
3. The glass shatters but remains in its frame or anchorage in a precarious condition, likely to fall out 
at any time. 
4. The glass falls out of its frame or anchorage, either in fragments, shards, or whole panels. 
Categories 1. and 2. provide both life safety and immediate occupancy levels of performance. In the 
case of category 2., even though the glass is cracked, it continues to provide a weather enclosure and 
barrier, and its replacement can be planned over a period of time. (Such glass replacement need not be 
performed in the immediate aftermath of the earthquake.) Categories 3. and 4. cannot provide for 
immediate occupancy, and their provision of a life safety level of performance depends on the postbreakage 
characteristics of the glass and the height from which it can fall. Tempered glass shatters into 
multiple, pebble-size fragments that fall from the frame or anchorage in clusters. These broken glass 
clusters are relatively harmless to humans when they fall from limited heights, but when they fall from 
greater heights they could be harmful. 
The requirement that .fallout not be less than 1.25IDP is derived from Earthquake Safety Design of 
Windows, published in November 1982 by the Sheet Glass Association of Japan. Eq. 6.3-1 is based on 
a similar equation in Bouwkamp and Meehan (1960) that permits calculation of the story drift required 
to cause glass-to-frame contact in a given rectangular window frame. Both calculations are based on the 
principle that a rectangular window frame (specifically, one that is anchored mechanically to adjacent 
stories of the primary structural system of the building) becomes a parallelogram as a result of story 
drift, and that glass-to-frame contact occurs when the length of the shorter diagonal of the parallelogram 
is equal to the diagonal of the glass panel itself. 
The 1.25 factor in the requirements described above reflect uncertainties associated with calculated 
inelastic seismic displacements in building structures. Wright (1989) stated that post-elastic 
deformations, calculated using the structural analysis process, may well underestimate the actual 
building deformation by up to 30 percent. It would therefore be reasonable to require the curtain wall 
glazing system to withstand 1.25 times the computed maximum interstory displacement to verify 
adequate performance. Therefore, Wrights comments form the basis for employing the 1.25 factor in 
these requirements. 
Introduction. Seismic design requirements for glass in building codes have traditionally been nonexistent 
or have been limited to the general statement that drift be accommodated. No distinction has 
been made regarding the seismic performance of different types of glass, different frames, and different 
glazing systems. Yet, significant differences exist in the performance of various glass types subjected to 
simulated earthquake conditions. Controlled laboratory studies were conducted to investigate the 
cracking resistance and fallout resistance of different types of glass installed in the same storefront and 
mid-rise wall systems. Effects of glass surface prestress, lamination, wall system type, and dry versus 
structural silicone glazing were considered. Laboratory results revealed that distinct magnitudes of 
story drift cause glass cracking and glass fallout in each glass type tested. Notable differences in 
seismic resistance exist between glass types commonly used in contemporary building design. 
Test rig and experimental plan. In-plane dynamic racking tests were performed using the rig shown in 
Figure C6.3-1. Rectangular steel tubes at the top and bottom of the facility are supported on roller 
assemblies, which permit only horizontal motion of the tubes. The bottom steel tube is driven by a 
Architectural, Mechanical, and Electrical Component Design Requirements 
123 
computer-controlled hydraulic ram, while the top tube is attached to the bottom tube by means of a 
fulcrum and pivot arm assembly. This mechanism causes the upper steel tube to displace the same 
amount as the lower steel tube, but in the opposite direction, which doubles the amount of story drift 
that can be imposed on a test specimen from " 76 mm (" 3 in.) to " 152 mm (" 6 in.). The test facility 
accommodated up to three glass test panels, each 1.5 m (5 ft) wide by 1.8 m (6 ft) high. A more detailed 
description of the dynamic racking test rig is included in Behr and Belarbi (1996). 
Figure C6.3-1 Dynamic racking test rig. 
Several types of glass, shown in Table C6.3-1, were tested under simulated seismic conditions in the 
storefront and mid-rise dynamic racking tests. These glass types, along with the wall systems employed 
in the tests, were selected after polling industry practitioners and wall system designers for their 
opinions regarding common glass types and common wall system types employed in contemporary 
storefront and mid-rise wall constructions. 
Storefront wall system tests. Tests were conducted on various glass types that were dry-glazed within 
a wall system, as commonly used in storefront applications. Loading histories for the storefront wall 
system tests were based on dynamic analyses performed on a typical storefront building that was not 
designed specifically for seismic resistance (Pantelides et al., 1996). Two types of tests were conducted 
on the storefront wall systems: (1) serviceability tests, wherein the drift loading history of the glass 
simulated the response of a storefront building structure to a maximum probable earthquake event; 
and (2) ultimate tests, wherein drift amplitudes were twice those of the serviceability tests, which was a 
simplified means of approximating the loading history of a maximum credible earthquake event. As 
indicated in Table C6.3-1, five glass types were tested, all dry-glazed in a storefront wall system. Three 
glass panels were mounted side by side in the test facility, after which horizontal (in-plane) racking 
motions were applied. 
Table C6.3-1 Glass Types Included in Storefront and Mid-Rise Dynamic Racking Tests 
Glass Type 
Storefront Tests 
Mid-Rise Tests 
6 mm (1/4 in.) annealed monolithic 
T 
T 
6 mm (1/4 in.) heat-strengthened monolithic 
T 
6 mm (1/4 in.) fully tempered monolithic 
T 
T 

2 003 Commentary, Chapter 6 
124 
6 mm (1/4 in.) annealed monolithic with 0.1 mm pet film (film 
not anchored to wall system frame) 
T 
6 mm (1/4 in.) annealed laminated 
T 
T 
6 mm (1/4 in.) heat-strengthened laminated 
T 
6 mm (1/4 in.) heat-strengthened monolithic spandrel 
T 
25 mm (1 in.) annealed insulating glass units 
T 
T 
25 mm (1 in.) heat-strengthened insulating glass units 
T 
The serviceability test lasted approximately 55 seconds and incorporated drift amplitudes ranging from 
" 6 to " 44 mm (" 0.25 to " 1.75 in.). The drift pattern in the ultimate test was formed by doubling 
each drift amplitude in the serviceability test. Both tests were performed at a nominal frequency of 0.8 
Hz. 
Experimental results indicated that for all glass types tested, serviceability limit states associated with 
glass edge damage and gasket seal degradation in the storefront wall system were exceeded during the 
moderate earthquake simulation (that is, the serviceability test). Ultimate limit states associated with 
major cracking and glass fallout were reached for the most common storefront glass type, 6 mm (1/4 in.) 
annealed monolithic glass, during the severe earthquake simulation (that is, the ultimate test). This 
observation is consistent with a reconnaissance report of damage resulting from the Northridge 
Earthquake (EERI, 1994). More information regarding the storefront wall system tests is included in 
Behr, Belarbi, and Brown (1995). In addition to the serviceability and ultimate tests, increasingamplitude 
crescendo tests, similar to those described below for the mid-rise tests, were performed at a 
frequency of 0.8 Hz on selected storefront glass types. Results of these crescendo tests are reported in 
Behr, Belarbi, and Brown (1995) and are included in some of the comparisons made below. 
Mid-rise curtain wall system tests. Another series of tests focused on the behavior of glass panels in a 
popular curtain wall system for mid-rise buildings. All mid-rise glass types in Table C6.3-1 were tested 
with a dry-glazed wall system that uses polymeric (rubber) gaskets wedged between the glass edges and 
the curtain wall frame to secure each glass panel perimeter. In addition, three glass types were tested 
with a bead of structural silicone sealant on the vertical glass edges and dry glazing gaskets on the 
horizontal edges (that is, a two-side structural silicone glazing system). Six specimens of each glass 
type were tested. 
Crescendo tests were performed on all mid-rise test specimens. As described by Behr and Belarbi 
(1996), the crescendo test consisted of a series of alternating ramp-up and constant amplitude 
intervals, each containing four sinusoidal-shaped drift cycles. Each drift amplitude step (that is, the 
increase in amplitude between adjacent constant amplitude intervals, which was achieved by completing 
the four cycles in the intermediary ramp-up interval) was " 6 mm (" 0.25 in.). The entire crescendo test 
sequence lasted approximately 230 seconds. Crescendo tests on mid-rise glass specimens were 
conducted at 1.0 Hz for dynamic racking amplitudes from 0 to 114 mm (0 to 4.5 in.), 0.8 Hz for 
amplitudes from 114 to 140 mm (4.5 to 5.5 in.), and 0.5 Hz for amplitudes from 140 to 152 mm (5.5 to 
6 in.). These frequency reductions at higher racking amplitudes were necessary to avoid exceeding the 
capacity of the hydraulic actuator ram in the dynamic racking test rig. 
The drift magnitude at which glass cracking was first observed was called the serviceability drift 
limit, which corresponds to the drift magnitude at which glass damage would necessitate glass 
replacement. The drift magnitude at which glass fallout occurred was called the ultimate drift limit, 
which corresponds to the drift magnitude at which glass damage would become a life safety hazard. 
This ultimate drift limit for architectural glass is related to .fallout in Sec. 6.3.7 of the Provisions, 
noting that horizontal racking displacements (drifts) in the crescendo tests were typically applied to test 
specimens having panel heights of only 1.8 m (6 ft). 
Architectural, Mechanical, and Electrical Component Design Requirements 
125 
In addition to recording the serviceability drift limit and ultimate drift limit for each glass test specimen, 
the drift magnitude causing first contact between the glass panel and the aluminum frame was also 
recorded. To establish when this contact occurred, thin copper wires were attached to each corner of the 
glass panel and were connected to an electronics box. If the copper wire came into contact with the 
aluminum frame, an indicator light on an electronics box was actuated. Measured drifts causing glassto-
aluminum contact correlated well with those predicted by Eq. 6.3-1. 
Glass failure patterns from crescendo tests. Glass failure patterns were recorded during each 
storefront test and mid-rise test. Annealed monolithic glass tended to fracture into sizable shards, which 
then fell from the curtain wall frame. Heat-strengthened monolithic glass generally broke into smaller 
shards than annealed monolithic glass, with the average shard size being inversely proportional to the 
magnitude of surface compressive prestress in the glass. Fully tempered monolithic glass shattered into 
much smaller, cube-shaped fragments. Annealed monolithic glass with unanchored 0.1 mm (4 mil) PET 
film also fractured into large shards, much like un-filmed annealed monolithic glass, but the shards 
adhered to the film. However, when the weight of the glass shards became excessive, the entire 
shard/film conglomeration sometimes fell from the glazing pocket as a unit. Thus, unanchored 0.1 mm 
PET film was not observed to be totally effective in terms of preventing glass fallout under simulated 
seismic loadings, which agrees with field observations made in the aftermath of the 1994 Northridge 
Earthquake (Gates and McGavin, 1998). Annealed and heat-strengthened laminated glass units 
experienced fracture on each glass ply separately, which permitted these laminated glass units to retain 
sufficient rigidity to remain in the glazing pocket after one (or even both), glass plies had fractured due 
to glass-to-aluminum contacts. Annealed and heat-strengthened laminated glass units exhibited the 
highest resistance to glass fallout during the dynamic racking tests. 
Quantitative drift limit data from crescendo tests. Serviceability and ultimate drift limit data 
obtained during the crescendo tests are presented in four panels in Figure C6.3-2. Figure C6.3-2a shows 
the effects of glass surface prestress (that is, annealed, heat-strengthened and fully tempered glass) on 
seismic drift limits; Figure C6.3-2b shows the effects of lamination (that is, monolithic glass, monolithic 
glass with unanchored 0.1 mm PET film, and laminated glass); Figure C6.-2c shows the effects of wall 
system type (that is, lighter, more flexible, storefront wall system versus the same glass types tested in a 
heavier, stiffer, mid-rise wall system); and Figure C6.3-2d shows the effects of structural silicone 
glazing (that is, dry glazing versus two-side structural silicone glazing). Each symbol plotted in Figure 
C6.3-2 is the mean value for specimens of a given glass type, along with " one standard deviation error 
bars. In those cases where error bars for a particular glass type overlap, only one side of the error bar is 
plotted. In cases where the glass panel did not experience fallout by the end of the crescendo test, a 
conservative ultimate drift limit magnitude of 152 mm (6 in.) (the racking limit of the test facility) is 
assigned for plotting purposes in Figure C6.3-2. (This ultimate drift limit, shown with a ? symbol in 
Figure C6.3-2, is related to the term .fallout in Sec. 6.3.7 of the Provisions.) No error bars are plotted 
for these pseudo data points, since the drift magnitude at which the glass panel would actually have 
experienced fallout could not be observed; certainly, the actual ultimate drift limits for these specimens 
are greater than "152 mm (" 6 in.). 
2 003 Commentary, Chapter 6 
126 
Figure C6.3-2 Seismic drift limits from crescendo tests on architectural glass 

Architectural, Mechanical, and Electrical Component Design Requirements 
127 
Figure C6.3-2 (continued) Seismic drift limits from crescendo tests on architectural glass 
The "152 mm (" 6 in.) racking limit of the test rig, when applied over the 1829 mm (72 in.) height of 
glazing panel specimens represents a severe story drift index of more than 8 percent. This 8 percent 
2 003 Commentary, Chapter 6 
128 
drift index exceeds, by a significant margin, provisions in Sec. 4.5.1 (Table 4.5-1) that set allowable 
drift limits between 0.7 percent and 2.5 percent, depending on structure type and Seismic Use Group. 
Thus, the drift limits, .a, in Table 4.5-1 are considerably lower than the racking limits of the laboratory 
facility used for the crescendo tests. In building design, however, values of .fallout would need to be 
significantly higher than the story drifts exhibited by the primary building structure in order to provide 
an acceptable safety margin against glass fallout. 
Summary observations from Figure C6.3-2. Effects of glass surface prestress - Figure C6.3-2a 
illustrates the effects of glass surface prestress on observed seismic drift limits. To eliminate all 
variables except for glass surface prestress, data from only the mid-rise curtain wall tests are plotted. 
Slight increases in cracking and fallout drift limits can be seen for 6 mm (0.25 in.) monolithic glass as 
the level of glass surface prestress increases from annealed to heat-strengthened to fully tempered glass. 
However, effects of glass surface prestress on observed seismic drift limits were statistically significant 
only when comparing 6 mm fully tempered monolithic glass to 6 mm annealed monolithic glass. All six 
of the 6 mm fully tempered monolithic glass specimens shattered when initial cracking occurred, 
causing the entire glass panels to fall out. Similar behavior was observed in four of the six 6 mm heatstrengthened 
monolithic glass specimens. No appreciable differences in seismic drift limits existed 
between annealed and heat-strengthened 25 mm (1 in.) insulating glass units. 
Effects of lamination -- Figure C6.3-2b shows the effects of lamination configuration on seismic drift 
limits. Lamination had no appreciable effect on the drift magnitudes associated with first observable 
glass cracking. In a dry-glazed system, the base glass type (and not the lamination configuration) 
appeared to control the drift magnitude associated with glass cracking. However, lamination 
configuration had a pronounced effect on glass fallout resistance (that is, .fallout). Specifically, 
monolithic glass types were more prone to glass fallout than were either annealed monolithic glass with 
unanchored 0.1 mm PET film or annealed laminated glass. All six annealed monolithic glass panels 
experienced glass fallout during the tests; five of six annealed monolithic glass specimens with 
unanchored 0.1 mm PET film experienced fallout; only one of six annealed laminated glass panels 
experienced fallout. 
Laboratory tests also showed that heat-strengthened laminated glass had higher fallout resistance than 
did heat-strengthened monolithic glass. Heat-strengthened monolithic glass panels fell out at 
significantly lower drift magnitudes than did heat-strengthened laminated glass units. Heatstrengthened 
laminated glass units tended to fall out in one large piece, instead of exhibiting the smaller 
shard fallout behavior of heat-strengthened monolithic glass. 
Effects of wall system type -- Figure C6.3-2c illustrates the effects of wall system type on observed 
seismic drift limits. For all four glass types tested in both the storefront and mid-rise wall systems, the 
lighter, more flexible storefront frames allowed larger drift magnitudes before glass cracking or glass 
fallout than did the heavier, stiffer mid-rise curtain wall frames. This observation held true for all glass 
types tested in both wall system types. 
Effects of structural silicone glazing -- As shown in Figure C6.3-2d, use of a two-side structural silicone 
glazing system increased the dynamic drift magnitudes associated with first observable glass cracking in 
both heat-strengthened monolithic glass and annealed insulating glass units. During the crescendo tests, 
glass panels were observed to walk horizontally across the frame after the beads of structural silicone 
sealant had sheared. Because the mid-rise curtain wall crescendo tests were performed on single glass 
panels, the glass specimen was unobstructed as it walked horizontally across the frame. In a multi-panel 
curtain wall assembly on an actual building, adjacent glass panels could collide, which could induce 
glass cracking at lower drift magnitudes than those observed in the single-panel tests performed in this 
study. It is also clear from Figure C6.3-2d that glass specimens with two-side structural silicone glazing 
exhibited higher resistance to glass fallout than did comparable dry-glazed glass specimens. 
Conclusion. Dynamic racking tests showed that distinct and repeatable dynamic drift magnitudes were 
associated with glass cracking and glass fallout in various types of glass tested in storefront and mid-rise 
Architectural, Mechanical, and Electrical Component Design Requirements 
129 
wall systems. Seismic resistance varied widely between glass types commonly employed in 
contemporary building design. Annealed and heat-strengthened laminated glass types exhibited higher 
resistance to glass fallout than did monolithic glass types. Annealed monolithic glass with unanchored 
0.1 mm PET film exhibited total fallout of the glass shard/adhesive film conglomeration in five out of 
six of the crescendo tests performed. 
Glass panels glazed within stiffer aluminum frames were less tolerant of glass-to-aluminum collisions 
and were associated with glass fallout events at lower drift magnitudes than were the same glass types 
tested in a more flexible aluminum frame. Glazing details were also found to have significant effects on 
the seismic performance of architectural glass. Specifically, architectural glass within a wall system 
using a structural silicone glaze on two sides exhibited higher seismic resistance than did identical glass 
specimens dry-glazed on all four sides within a comparable wall framing system. 
6.4 MECHANICAL AND ELECTRICAL COMPONENTS 
The primary focus of these requirements is on the design of attachments and equipment supports for 
mechanical and electrical components. 
The requirements are intended to reduce the hazard to life posed by the loss of component structural 
stability or integrity. The requirements should increase the reliability of component operation but do 
not directly address the assurance of functionality. For critical components where operability is vital, the 
requirements of Sec. 2.4.5 provide methods for seismically qualifying the component. 
The design of mechanical and electrical components must consider two levels of earthquake safety. For 
the first safety level, failure of the mechanical or electrical component itself poses no significant hazard. 
In this case, the only hazard posed by the component is if the support and the means by which the 
component and its supports are attached to the building or the ground fails and the component could 
slide, topple, fall, or otherwise move in a manner that creates a hazard for persons nearby. In the first 
category, the intent of these requirements is only to design the support and the means by which the 
component is attached to the structure, defined in the Definitions as supports and attachments. For 
the second safety level, failure of the mechanical or electrical equipment itself poses a significant 
hazard. This could be a case either of failure of a containment having hazardous contents or contents 
required after the earthquake or of functional failure of a component required to remain operable after 
an earthquake. In this second category, the intent of these requirements is to provide guidance for the 
design of the component as well as the means by which the component is supported and attached to the 
structure. The requirements should increase the survivability of this second category of component but 
the assurance of functionality may require additional considerations. 
Examples of this second category include fire protection piping or an uninterruptible power supply in a 
hospital. Another example involves the rupture of a vessel or piping that contains sufficient quantities 
of highly toxic or explosive substances such that a release would be hazardous to the safety of building 
occupants or the general public. In assessing whether failure of the mechanical or electrical equipment 
itself poses a hazard, certain judgments may be necessary. For example, small, flat-bottom tanks 
themselves may not need to be designed for earthquake loads, but the hazard of a large fluid spill 
associated with seismic failure of large, flat-bottom tanks suggests that the design of many, if not most, 
of such tanks should consider earthquake loads. Distinguishing between large and small, in this case, 
may require an assessment of potential damage caused by a spill of the fluid contents as outlined in Sec. 
6.2.3 and Chapter 1 of ASCE-7 
It is intended that the requirements provide guidance for the design of components for both conditions 
in the second category. This is primarily accomplished by increasing the design forces with an 
importance factor, Ip. However, this directly affects only structural integrity and stability. Function and 
operability of mechanical and electrical components may be affected only indirectly by increasing 
design forces. For complex components, testing or experience may be the only reasonable way to 
improve the assurance of function and operability. On the basis of past earthquake experience, it may 
2 003 Commentary, Chapter 6 
130 
be reasonable to conclude that if structural integrity and stability are maintained, function and 
operability after an earthquake will be provided for many types of equipment components. On the other 
hand, mechanical joints in containment components (tanks, vessels, piping, etc.) may not remain 
leaktight in an earthquake even if leaktightness is re-established after the earthquake. Judgment may 
suggest a more conservative design related in some manner to the perceived hazard than would 
otherwise be provided by these requirements. 
It is not intended that all equipment or parts of equipment be designed for seismic forces. 
Determination of whether these requirements need to be applied to the design of a specific piece of 
equipment or a part of that equipment will sometimes be a difficult task. When Ip is specified to be 1.0, 
damage to, or even failure of, a piece or part of a component is not a concern of these requirements so 
long as a hazard to life does not exist. Therefore, the restraint or containment of a falling, breaking, or 
toppling component (or its parts) by the use of bumpers, braces, guys, wedges, shims, tethers, or gapped 
restraints often may be an acceptable approach to satisfying these requirements even though the 
component itself may suffer damage. Judgment will be required if the intent of these requirements is to 
be fulfilled. The following example may be helpful: Since the threat to life is a key consideration, it 
should be clear that a nonessential air handler package unit that is less than 4 ft (1.2 m) tall bolted to a 
mechanical room floor is not a threat to life as long as it is prevented from significant motions by having 
adequate anchorage. Therefore, earthquake design of the air handler itself need not be performed. 
However, most engineers would agree that a 10-ft (3.0 m) tall tank on 6-ft (1.8 m) angles used as legs 
mounted on the roof near a building exit does pose a hazard. It is the intent of these requirements that 
the tank legs, the connections between the roof and the legs, the connections between the legs and the 
tank, and possibly even the tank itself be designed to resist earthquake forces. Alternatively, restraint of 
the tank by guys or bracing could be acceptable. Certain suspended components are exempt from lateral 
bracing requirements, provided they meet prescriptive force and interaction requirements. 
It is not the intent of the Provisions to require the seismic design of shafts, buckets, cranks, pistons, 
plungers, impellers, rotors, stators, bearings, switches, gears, nonpressure retaining casings and castings, 
or similar items. Where the potential for a hazard to life exists, it is expected that design efforts will 
focus on equipment supports including base plates, anchorages, support lugs, legs, feet, saddles, skirts, 
hangers, braces, or ties. 
Many mechanical and electrical components consist of complex assemblies of mechanical and/or 
electrical parts that typically are manufactured in an industrial process that produces similar or identical 
items. Such equipment may include manufacturer's catalog items and often are designed by empirical 
(trial-and-error) means for functional and transportation loadings. A characteristic of such equipment is 
that it may be inherently rugged. Rugged, as used herein, refers to an ampleness of construction that 
provides such equipment with the ability to survive strong motions without significant loss of function. 
By examining such equipment, an experienced design professional usually should be able to confirm 
such ruggedness. The results of an assessment of equipment ruggedness will b used in determining an 
appropriate method and extent of seismic design or qualification efforts. 
It also is recognized that a number of professional and industrial organizations have developed 
nationally recognized codes and standards for the design and construction of specific mechanical and 
electrical components. In addition to providing design guidance for normal and upset operating 
conditions and various environmental conditions, some have developed earthquake design guidance in 
the context of the overall mechanical or electrical design. It is the intent of these requirements that such 
codes and standards having earthquake design guidance be used; normally the developers of such codes 
and standards are more familiar with the expected failure modes of the components for which they have 
developed design and construction rules. In particular, such codes and standards may be based on 
considerations that are not immediately obvious to the structural design professional. For example, in 
the design of industrial piping, seismic loads are not typically additive to thermal expansion loads. 
Given the potential for misunderstanding and mis-application of codes and standards specific to the 
design of mechanical and electrical systems, it is recommended that a registered design professional 
Architectural, Mechanical, and Electrical Component Design Requirements 
131 
familiar with these Provisions, as well as the those of the referenced code or standard used to evaluate 
the capacity of the mechanical or electrical components, should be involved in the review and 
acceptance of the seismic design process for such components. In addition, even if such codes and 
standards do not have earthquake design guidance, it is generally regarded that construction of 
mechanical and electrical equipment to nationally recognized codes and standards such as those 
approved by the American National Standards Institute provide adequate strength (with a safety margin 
often greater than that provided by structural codes) to accommodate all normal and upset operating 
loads. In this case, it could also be assumed that the component (especially if constructed of ductile 
materials) will not break up or break away from its supports in such a way as to pose a life-safety 
hazard. Earthquake damage surveys confirm this. 
Specific guidance for selected components or conditions is provided in Sec. 6.4.5 through 6.4.9. 
Testing is a well established alternative method of seismic qualification for small to medium size 
equipment. Several national standards have testing requirements adaptable for seismic qualification. 
AC 156, Acceptance Criteria for Seismic Qualification Testing of Nonstructural Components, is 
an acceptable shake table testing protocol, which meets the force requirements of the Provisions as well 
as ASCE 7-02. 
6.4.1 Component period. Determination of the fundamental period of an item of mechanical or 
electrical equipment using analytical or in-situ testing methods can become very involved and can 
produce nonconservative results (that is, underestimated fundamental periods) if not properly 
performed. 
When using analytical methods, it is absolutely essential to define in detail the flexibility of the 
elements of the equipment base, load path, and attachment to determine Kp. This base flexibility 
typically dominates equipment component flexibility and thus fundamental period. 
When using test methods, it is necessary to ensure that the dominant mode of vibration of concern for 
seismic evaluation is excited and captured by the testing. This dominant mode of vibration typically 
cannot be discovered in equipment in-situ tests that measure only ambient vibrations. In order to excite 
the mode of vibration with the highest fundamental period by in-situ tests, relatively significant input 
levels of motion are required (that is, the flexibility of the base and attachment needs to be exercised). 
A procedure such as the resonant frequency search in AC 156 may used to identify the dominant modes 
of vibration of the component. 
Many types of mechanical components have fundamental periods below 0.06 sec and may be 
considered to be rigid. Examples include horizontal pumps, engine generators, motor generators, air 
compressors, and motor driven centrifugal blowers. Other types of mechanical equipment also are very 
stiff but may have fundamental periods up to approximately 0.125 sec. Examples of these mechanical 
equipment items include vertical immersion and deep well pumps, belt driven and vane axial fans, 
heaters, air handlers, chillers, boilers, heat exchangers, filters, and evaporators. These fundamental 
period estimates do not apply when the equipment is on vibration-isolator supports. 
Electrical equipment cabinets can have fundamental periods of approximately 0.06 to 0.3 sec depending 
upon weight, stiffness of the enclosure assembly, flexibility of the enclosure base, and load path through 
to the attachment points. Tall, narrow motor control centers and switchboards lie in the upper end of 
this period range. Low- and medium-voltage switchgear, transformers, battery chargers, inverters, 
instrumentation cabinets, and instrumentation racks usually have fundamental periods ranging from 0.1 
to 0.2 sec. Braced battery racks, stiffened vertical control panels, benchboards, electrical cabinets with 
top bracing, and wall-mounted panelboards have fundamental periods ranging from 0.06 to 0.1 sec. 
6.4.2 and 6.4.3 Mechanical and electrical components. Past earthquakes have demonstrated that 
most mechanical and electrical equipment is inherently rugged and performs well provided that it is 
properly attached to the structure. This is because the operational and transportation loads for which the 
equipment is designed are typically larger than those due to earthquakes. For this reason, the 
2 003 Commentary, Chapter 6 
132 
requirements focus primarily on equipment anchorage and attachments. However, it was felt that 
mechanical components required to maintain containment of flammable or hazardous materials should 
themselves be designed for seismic forces. 
In addition, the reliability of equipment operability after an earthquake can be increased if the following 
items are also considered in design: 
1. Internal assemblies, subassemblies, and electrical contacts are attached sufficiently to prevent their 
being subjected to differential movement or impact with other internal assemblies or the equipment 
enclosure. 
2. Operators, motors, generators, and other such components that are functionally attached to 
mechanical equipment by means of an operating shaft or mechanism are structurally connected or 
commonly supported with sufficient rigidity such that binding of the operating shaft will be 
avoided. 
3. Any ceramic or other nonductile components in the seismic load path are specifically evaluated. 
4. Adjacent electrical cabinets are bolted together and cabinet lineups are prevented from banging into 
adjacent structural members. 
Components that could be damaged or could damage other components and are fastened to multiple 
locations of a structure should be designed to accommodate seismic relative displacements. Examples 
of components that should be designed to accommodate seismic relative displacements include bus 
ducts, cable trays, conduit, elevator guide rails, and piping systems. 
6.4.4 Supports and attachments. For some items such as piping, relative seismic displacements 
between support points generally are of more significance than inertial forces. Components made of 
ductile materials such as steel or copper can accommodate relative displacement effects by inelastically 
conforming to the support conditions. However, components made of less ductile materials can only 
accommodate relative displacement effects if appropriate flexibility or flexible connections are 
provided. 
Of most concern are distribution systems that are a significant life-safety hazard and are routed between 
two separate building structures. Ductile components with bends and elbows at the building separation 
point or components that will be subject to bending stresses rather than direct tensile loads due to 
differential support motion are less prone to damage and are less likely to fracture and fall provided the 
supports can accommodate the imposed loads. 
It is the intent of these requirements to ensure that all mechanical and electrical component supports be 
designed to accommodate the force and displacement effects prescribed. Component supports are 
differentiated here from component attachments to emphasize that the supports themselves, the 
structural members, braces, frames, skirts, legs, saddles, pedestals, cables, guys, stays, snubbers, and 
tethers, even if fabricated with and/or by the mechanical or electrical component manufacturer, should 
be designed for seismic forces. This is regardless of whether the mechanical or electrical component 
itself is designed for seismic loads. The intention is to prevent a component from sliding, falling, 
toppling, or otherwise moving such that the component would imperil life. 
6.4.5 Utility and service lines. For essential facilities, auxiliary on-site mechanical and electrical 
utility sources are recommended. It is recommended that an appropriate clause be included if existing 
codes for the jurisdiction do not presently provide for it. 
Sec. 6.4.5 requires that adequate flexibility be provided for utilities at the interface of adjacent and 
independent structures to accommodate anticipated differential displacement. It affects architectural and 
mechanical/electrical fittings only where water and energy lines pass through the interface. The 
displacements considered must include the Cd factor of Sec. 4.3.1 and should be in accordance with 
Provisions Sec. 6.2.7. 
Architectural, Mechanical, and Electrical Component Design Requirements 
133 
Consideration may be necessary for nonessential piping that carries quantities of materials that could 
damage essential utilities in the event of pipe rupture. 
Following a review of information from the Northridge and Loma Prieta earthquakes and discussions 
with gas company personnel, automatic earthquake shutoff of gas lines at structure entry points is no 
longer required. The primary justification for this is the consensus opinion that shutoff devices tend to 
cause more problems than they solve. Commercially available shutoff devices are often susceptible to 
inadvertent shutoff caused by passing vehicles and other non-seismic vibrations. This leads to 
disruption of service and often requires that local gas companies reset such devices and relight any pilot 
lights. In an earthquake, the majority of shutoff devices which actuate will be attached to undamaged 
gas lines. This results in a huge relight effort for the local utility at a time when resources are typically 
at a premium. If the earthquake occurs during the winter, a greater life hazard may exist from a lack of 
gas supply than from potential gas leaks. In the future, as shutoff devices improve and gas-fired 
appliances which use pilots are phased out, it may be justified to require shutoff devices. 
This is not meant to discourage individuals and companies from installing shutoff devices. In particular, 
individuals and companies who are capable of relighting gas-fired equipment should seriously consider 
installation of these devices. In addition, gas valves should be closed whenever leaks are detected. 
6.4.6 HVAC ductwork. Experience in past earthquakes has shown that, in general, HVAC duct 
systems are rugged and perform well in strong shaking motions. Bracing in accordance with the Sheet 
Metal and Air Conditioning Contractors National Association SMACNA 80, SMACNA 95, and 
SMACNA 98 has been shown to be effective in limiting damage to duct systems under earthquake 
loads. Typical failures have affected system function only and major damage or collapse has been 
uncommon. Therefore, industry standard practices should prove adequate for most installations. 
Expected earthquake damage should be limited to opening of the duct joints and tears in the ducts. 
Connection details that are prone to brittle failures, especially hanger rods subject to large amplitude 
cycles of bending stress, should be avoided. 
Some ductwork systems carry hazardous materials or must remain operational during and after an 
earthquake. Such ductwork system would be assigned a value of Ip greater than 1.0. A detailed 
engineering analysis for these systems should be performed. 
All equipment attached to the ducts and weighing more than 75 lb (334 N) , such as fans, humidifiers, 
and heat exchangers, should be braced independently of the duct. Unbraced in-line equipment can 
damage the duct by swinging and impacting it during an earthquake. Items attached to the duct (such as 
dampers, louvers, and air diffusers) should be positively supported by mechanical fasteners (not 
friction-type connections) to prevent their falling during an earthquake. 
Where it is desirable to limit the deflection of duct systems under seismic load, bracing in accordance 
with the SMACNA references listed in Sec. 6.1.1 may be used. 
6.4.7 Piping systems. Experience in past earthquakes has shown that, in general, piping systems are 
rugged and perform well in strong shaking motions. Numerous standards and guidelines have been 
developed covering a wide variety of piping systems and materials. Construction in accordance with 
current requirements of the referenced national standards have been shown to be effective in limiting 
damage to and avoiding loss of fluid containment in piping systems under earthquake conditions. It is 
therefore the intention of the Provisions that nationally recognized standards be used to design piping 
systems provided that the force and displacement demands are equal to or exceed those outlined in Sec. 
6.2.6 and 6.2.7 and provision is made to mitigate seismic interaction issues not normally addressed in 
the national standards. 
The following industry standards, while not adopted by ANSI, are in common use and may be 
appropriate reference documents for use in the seismic design of piping systems: SMACNA Guidelines 
for the Seismic Restraint of Mechanical Systems and ASHRAE CH 50-95 Seismic Restraint Design 
Piping. 
2 003 Commentary, Chapter 6 
134 
6.4.8 Boilers and pressure vessels. Experience in past earthquakes has shown that, in general, boilers 
and pressure vessels are rugged and perform well in strong shaking motions. Construction in 
accordance with current requirements of the ASME Boiler and Pressure Vessel Code (ASME BPV) has 
been shown to be effective in limiting damage to and avoiding loss of fluid containment in boilers and 
pressure vessels under earthquake conditions. It is therefore the intention of the Provisions that 
nationally recognized codes be used to design boilers and pressure vessels provided that the seismic 
force and displacement demands are equal to or exceed those outlined in Sec. 6.2.6 and 6.2.7. Until 
such nationally recognized codes incorporate force and displacement requirements comparable to the 
requirements of Sec. 6.2.6 and 6.2.7, it is nonetheless the intention to use the design acceptance criteria 
and construction practices of those codes. 
6.4.9 Elevators. The ASME Safety Code for Elevators and Escalators (ASME A17.1) has adopted 
many requirements to improve the seismic response of elevators; however, they do not apply to some 
regions covered by this chapter. These changes are to extend force requirements for elevators to be 
consistent with the Provisions. 
6.4.9.2 Elevator machinery and controller supports and attachments. ASME A17.1 has no seismic 
requirements for supports and attachments for some structures and zones where the Provisions are 
applicable. Criteria are provided to extend force requirements for elevators to be consistent with the 
intent and scope of the Provisions. 
6.4.9.3 Seismic switches. The purpose of seismic switches as used here is different from that of 
ASME A17.1, which has incorporated several requirements to improve the seismic response of elevators 
(such as rope snag point guard, rope retainer guards, and guide rail brackets) and which does not apply 
to some buildings and zones covered by the Provisions. Building motions that are expected in these 
uncovered seismic zones are sufficiently large to impair the operation of elevators. The seismic switch 
is positioned high in the structure where structural response will be the most severe. The seismic switch 
trigger level is set to shut down the elevator when structural motions are expected to impair elevator 
operations. 
Elevators in which the seismic switch and counterweight derail device have triggered should not be put 
back into service without a complete inspection. However, in the case where the loss of use of the 
elevator creates a life-safety hazard, an attempt to put the elevator back into service may be attempted. 
Operating the elevator prior to inspection may cause severe damage to the elevator or its components. 
The building owner should have detailed written procedures in place defining for the elevator 
operator/maintenance personnel which elevators in the facility are necessary from a post-earthquake life 
safety perspective. It is highly recommended that these procedures be in-place, with appropriate 
personnel training, prior to an event strong enough to trip the seismic switch. 
Once the elevator seismic switch is reset, it will respond to any call at any floor. It is important that the 
detailed procedure include the posting of out-of-service for testing signs at each door at each floor, 
prior to resetting the switch. Once the testing is completed and the elevator operator/maintenance 
personnel are satisfied that the elevator is safe to operate, the signs can be removed. 
6.4.9.4 Retainer plates. The use of retainer plates is a very low cost provision to improve the seismic 
response of elevators. 
Architectural, Mechanical, and Electrical Component Design Requirements 
135 
RELATED CONCERNS 
Maintenance. Mechanical and electrical devices installed to satisfy the requirements of the Provisions 
(for example, resilient mounting components or certain protecting devices) require maintenance to 
ensure their reliability and to provide protection in case of a seismic event for which they are designed. 
Specifically, rubber-in-shear mounts or spring mounts (if exposed to weathering) may deteriorate with 
time and, thus, periodic testing is required to ensure that their damping action will be available during 
an earthquake. Pneumatic mounting devices and electric switchgear must be maintained free of dirt and 
corrosion. How a regulatory agency could administer such periodic inspections has not been 
determined, so requirements to cover these situations have not been included. 
Tenant improvements. It is intended that the requirements in Chapter 6 also apply to newly 
constructed tenant improvements that are listed in Tables 6.3-1 and 6.4-1 and that are installed at any 
time during the life of the structure. 
Minimum standards. Criteria represented in the Provisions represent minimum standards. They are 
designed to minimize hazard for occupants and to improve the likelihood of functioning of facilities 
required by the community to deal with the consequences of a disaster. They are not designed to protect 
the owner's investment, and the designer of the facility should review with the owner the possibility of 
exceeding these minimum standards so as to limit his economic risk. 
The risk is particularly acute in the case of sealed, air-conditioned structures where downtime after a 
disaster can be materially affected by the availability of parts and labor. The parts availability may be 
significantly worse than normal because of a sudden increase in demand. Skilled labor also may be in 
short demand since available labor forces may be diverted to high priority structures requiring repairs. 
Architect-Engineer design integration. The subject of architect-engineer design integration is raised 
here because it is believed that all members of the profession should clearly understand that Chapter 6 is 
a compromise based on concerns for enforcement and the need to develop a simple, straightforward 
approach. It is imperative that, from the outset, architectural input concerning definition of occupancy 
classification and the required level of seismic resistance be properly considered in the structural 
engineers approach to seismic safety if the design profession as a whole is to make any meaningful 
impact on the public awareness of this matter. It is hoped that, as the design profession gains more 
knowledge and sophistication in the use of seismic design, a more comprehensive approach to 
earthquake design requirements will be developed. 
REFERENCES 
ANCO Engineers, Inc. 1983. Seismic Hazard Assessment of Non-Structural ComponentsPhase I, 
Final Report for National Science Foundation from ANCO Engineers, Inc., Culver City, California, 
September. 
Bachman, R. E, and R. M. Drake. 1996. A Study To Empirically Validate the Component Response 
Modification Factors in the 1994 NEHRP Provisions Design Force Equations for Architectural, 
Mechanical, and Electrical Components, letter report to the National Center for Earthquake 
Engineering Research, July. 
Bachman, R. E., R. M. Drake, and P. J. Richter. 1993. 1994 Update to 1991 NEHRP Provisions for 
Architectural, Mechanical, and Electrical Components and Systems, letter report to the National Center 
for Earthquake Engineering Research, February 22. 
Behr, R.A., Belarbi, A. and Brown, A.T., 1995, Seismic Performance of Architectural Glass in a 
Storefront Wall System, Earthquake Spectra, Vol. 11, No. 3, pp. 367-391. 
Behr, R.A. and Belarbi, A., 1996, Seismic Tests Methods for Architectural Glazing Systems, 
Earthquake Spectra, Vol. 12, No. 1, pp. 129-143. 
2 003 Commentary, Chapter 6 
136 
Bouwkamp, J. G. and Meehan, J. F. 1960. Drift Limitations Imposed by Glass, Proceedings of the 
Second World Conference on Earthquake Engineering. Tokyo, Japan. pp. 1763-1778. 
Drake, R. M., and R. E. Bachman. 1996. NEHRP Provisions for 1994 for Nonstructural 
Components, ASCE Journal of Architectural Engineering, March. 
Drake, R. M., and R. E. Bachman. 1995. Interpretation of Instrumented Building Seismic Data and 
Implications for Building Codes, in Proceedings of the 1995 SEAOC Annual Convention. 
Drake, R. M., and R. E. Bachman. 1994. 1994 NEHRP Provisions for Architectural, Mechanical, and 
Electrical Components, in Proceedings, 5th United States National Conference on Earthquake 
Engineering. 
Earthquake Engineering Research Institute (EERI), 1994, Northridge Earthquake January 17, 1994: 
Preliminary Reconnaissance Report, ed. John F. Hall, Oakland, California, pp. 56-57. 
Gates, W.E. and McGavin, G., 1998, Lessons Learned from the 1994 Northridge Earthquake on the 
Vulnerability of Nonstructural Systems, Proceedings, Seminar on Seismic Design, Retrofit, and 
Performance of Nonstructural Components, ATC-29-1, January 22-23, 1998, San Francisco, California, 
Applied Technology Council, Redwood City, California, pp. 93-101. 
Housner, G. W., and M. A. Haroun. 1980. Seismic Design of Liquid Storage Tanks in ASCE 
Convention Proceedings. 
Pantelides, C.P., Truman, K.Z., Behr, R.A. and Belarbi, A., 1996, Development of a Loading History 
for Seismic Testing of Architectural Glass in a Shop-Front Wall System, Engineering Structures, Vol. 
18, No. 12, pp. 917-935. 
Wright, P. D. 1989. The Development of a Procedure and Rig for Testing the Racking Resistance of 
Curtain Wall Glazing; Building Research Association of New Zealand (BRANZ) Study Report No. 17; 
p. 18. 
Acknowledgment 
The National Science Foundation (Grant No. CMS 9213172) provided major funding for the 
experimental results summarized in Sec. C6.3.7. 
End note: The American Architectural Manufacturers Association (AAMA) has issued 
AAMA 501.4-2000: Recommended Static Test Method for Evaluating Curtain Wall and Storefront 
Systems Subjected to Seismic and Wind Induced Interstory Drifts. In contrast with the dynamic 
displacements employed in the crescendo tests described in this section, static displacements are 
employed in AAMAs recommended test method. Correlations between the results of the static and 
dynamic test methods have not yet been established with regard to the seismic performance of 
architectural glazing systems. 
Note that this section addresses glass in frames subject to interstory drift. Glass frames not subject to 
interstory drift, such as freestanding guards and screens, are not covered by this section. 

Architectural, Mechanical, and Electrical Component Design Requirements 
137 
This page is intentionally left blank. 
 137 
Appendix to Chapter 6 Commentary 
ALTERNATIVE PROVISIONS FOR THE DESIGN OF PIPING 
SYSTEMS 
A6.1 Seismic Interaction. There are two types of seismic interactions: system interactions and spatial 
interactions. A system interaction is a spurious or erroneous signal resulting in unanticipated operating 
conditions, such as the spurious start-up of a pump motor or the unintended closure of a valve. Spatial 
interactions are interactions caused by the failure of a structure or component in close proximity. Spatial 
interactions can in turn be further divided into falling interactions, swing interactions, and spray 
interactions. A falling interaction is an impact on a critical component due to the fall of overhead or 
adjacent equipment or structure. A swing interaction is an impact due to the swing or rocking of 
adjacent component or suspended system. A spray interaction is due to the leakage of overhead or 
adjacent piping or vessels. 
Any interaction involves two components, a source and a target. An interaction source is the component 
or structure that could fail and interact with the seismically designed component. An interaction target is 
a seismically qualified component that is being impacted, sprayed or spuriously activated. For an 
interaction to affect a seismically qualified component, it must be credible and significant. A credible 
interaction is one that can take place. For example, the fall of a ceiling panel located overhead from a 
motor control center is a credible interaction because the falling panel can reach and impact the motor 
control center. The target (the MCC) is said to be within the zone of influence of the source (the ceiling 
panel). A significant interaction is one that can result in damage to the target. For example, the fall of a 
light fixture on a 20 steel pipe may be credible (the light fixture being above the pipe) but may not be 
significant (the light fixture will not damage the steel pipe). In contrast, the overturning of a rack on an 
instrument panel is a significant interaction. 
The process of considering seismic interactions begins with a interaction review. For new structures, 
this involves examination of the design drawings, to identify the interaction targets, and credible and 
significant sources of interaction. In many cases, the design documents may only locate components and 
systems in schematic terms. The actual location of, for example, piping and ductwork systems is 
determined in the field. In this case, and where work is being performed on an existing structure, it is 
necessary to begin the interaction review with a walk-down, typically with a photographic record. 
Based on the assembled data, supporting calculations to document credible and significant sources of 
interactions can be prepared. 
In practice, it is only necessary to document credible and significant sources of interaction. It is not 
necessary to list and evaluate every single overhead or adjacent component in the area around the target, 
only those that could interact and whose interaction could damage the target. Because only credible and 
significant sources of interaction are documented, an important aspect of the interaction review is 
engineering judgment. The spatial interaction review should therefore be performed by experienced 
seismic design engineers. 
Where system interactions are of importance the written input of a system engineer is in order. 
2003 Commentary, Appendix to Chapter 6 
138 
This page is intentionally left blank. 

 139 
Chapter 7 Commentary 
FOUNDATION DESIGN REQUIREMENTS 
7.1 GENERAL 
7.1.1 Scope. The minimum foundation design requirements that might be suitable when any 
consideration must be given to earthquake resistance are set forth in Chapter 7. It is difficult to separate 
foundation requirements for minimal earthquake resistance from the requirements for resisting normal 
vertical loads. In order to have a minimum base from which to start, this chapter assumes compliance 
with all basic requirements necessary to provide support for vertical loads and lateral loads other than 
earthquake. These basic requirements include, but are not limited to, provisions for the extent of 
investigation needed to establish criteria for fills, slope stability, expansive soils, allowable soil 
pressures, footings for specialized construction, drainage, settlement control, and pile requirements and 
capacities. Certain detailing requirements and the allowable stresses to be used are provided in other 
chapters of the Provisions as are the additional requirements to be used in more seismically active 
locations. 
7.2 GENERAL DESIGN REQUIREMENTS 
7.2.2 Soil capacities. This section requires that the building foundation without seismic forces applied 
must be adequate to support the building gravity load. When seismic effects are considered, the soil 
capacities can be increased considering the short time of loading and the dynamic properties of the soil. 
It is noted that the Appendix to Chapter 7 introduces into the Provisions ultimate strength design (USD) 
procedures for the geotechnical design of foundations. The Commentary Appendix to Chapter 7 
provides additional guidance and discussion of the USD procedures. 
7.2.3 Foundation load-deformation characteristics. The Appendix to Chapter 7 (Provisions and 
Commentary (Sec. A7.2.3) provides guidance on modeling load-deformation characteristics of the 
foundation-soil system (foundation stiffness). The guidance contained therein covers both linear and 
nonlinear analysis methods. 
7.3 SEISMIC DESIGN CATEGORY B 
There are no special seismic provisions for the design of foundations for buildings assigned to Seismic 
Design Category B. 
7.4 SEISMIC DESIGN CATEGORY C 
Extra precautions are required for the seismic design of foundations for buildings assigned to Seismic 
Design Category C. 
7.4.1 Investigation. This section reviews procedures that are commonly used for evaluating potential 
site geologic hazards due to earthquakes, including slope instability, liquefaction, and surface fault 
rupture. Geologic hazards evaluations should be carried out by qualified geotechnical professionals and 
documented in a written report. 
Screening Evaluation. Evaluation of geologic hazard may initially consist of a screening evaluation. 
If the screening evaluation clearly demonstrates that a hazard is not present, then more detailed 
evaluations, using procedures such as those described in the following sections, need not be conducted. 
Reference to the following publications are suggested for guidelines on screening evaluations: 
California Division of Mines and Geology (1997)  slope instability; Blake et al. (2002) and Stewart et 
al. (2003)  slope instability; Martin and Lew (1999)  liquefaction; U.S. Army Corps of Engineers 
(1998)  slope instability; liquefaction; surface fault rupture. More detailed evaluation procedures such 
as those described below should be used if a hazard cannot be screened out. 
2003 Commentary, Chapter 7 
140 
Slope instability hazard. The stability of slopes composed of dense (nonliquefiable) or nonsaturated 
sandy soils or nonsensitive clayey soils can be determined using standard procedures. 
For initial evaluation, the pseudostatic analysis may be used. (The deformational analysis described 
below, however, is now preferred.) In the pseudostatic analysis, inertial forces generated by earthquake 
shaking are represented by an equivalent static horizontal force acting on the slope. The seismic 
coefficient for this analysis should be the peak ground acceleration, amax or SDS/2.5. The factor of safety 
for a given seismic coefficient can be estimated by using traditional slope stability calculation methods. 
A factor of safety greater than one indicates that the slope is stable for the given lateral force level and 
further analysis is not required. A factor of safety of less than one indicates that the slope will yield and 
slope deformation can be expected and a deformational analysis should be made using the techniques 
discussed below. 
Deformational analyses resulting in estimates of slope displacement are now accepted practice. The 
most common analysis, termed a Newmark analysis (Newmark, 1965), uses the concept of a frictional 
block sliding on a sloping plane or arc. In this analysis, seismic inertial forces are calculated using a 
time history of horizontal acceleration as the input motion. Slope movement occurs when the driving 
forces (gravitational plus inertial) exceed the resisting forces. This approach estimates the cumulative 
displacement of the sliding mass by integrating increments of movement that occur during periods of 
time when the driving forces exceed the resisting forces. Displacement or yield occurs when the 
earthquake ground accelerations exceed the acceleration required to initiate slope movement or yield 
acceleration. The yield acceleration depends primarily on the strength of the soil and the gradient and 
height and other geometric attributes of the slope. See Figure C7.4-1 for forces and equations used in 
analysis and Figure C7.4-2 for a schematic illustration for a calculation of the displacement of a soil 
block toward a bluff. 
Acceptable methods for the determination of displacements on many projects involve the use of charts 
that show displacements for different acceleration ratios, where the acceleration ratio is defined as the 
ratio of yield acceleration to peak ground acceleration. Various charts have been developed, including 
those by Franklin and Chang (1977), Makdisi and Seed (1978), Wong and Whitman (1982), Hynes and 
Franklin (1984), Martin and Qiu (1994); and Bray and Rathje (1998). The selection between the 
different charts should be made on the basis of the type of slope and the degree of conservatism 
necessary for the project. A number of the chart methods were developed for the estimation of 
displacements for dams, and therefore, may be more suitable for embankment designs. 
Recommendations on the use of such procedures for typical building construction are presented by 
Blake et al. (2002). 
Foundation Design Requirements 
141 
Mitigation of slope instability hazard. With respect to slope instability, three general mitigative 
measures might be considered: design the structure to resist the hazard, stabilize the site to reduce the 
hazard, or choose an alternative site. Ground displacements generated by slope instability are similar in 
destructive character to fault displacements generating similar senses of movement: compression, shear, 
extension or vertical. Thus, the general comments on structural design to prevent damage given under 
mitigation of fault displacement apply equally to slope displacement. Techniques to stabilize a site 
include reducing the driving forces by grading and drainage of slopes and increasing the resisting forces 
by subsurface drainage, buttresses, ground anchors, reaction piles or shafts, ground improvement using 
densification or soil mixing methods, or chemical treatment. 
2003 Commentary, Chapter 7 
142 
Liquefaction hazard. Liquefaction of saturated granular soils has been a major source of building 
damage during past earthquakes. Loss of bearing strength, differential settlement, and horizontal 
displacement due to lateral spreads have been the direct causes of damage. Examples of this damage 
can be found in reports from many of the more recent earthquakes in the United States, including the 
1964 Alaska, the 1971 San Fernando, the 1989 Loma Prieta, the 1994 Northridge, the 2001 Nisqually, 
and the 2003 Denali earthquakes. Similar damage was reported after the 1964 Niigata, the 1994 
Hyogoken-Nanbu (Kobe), the 1999 Taiwan, and the 1999 Turkey earthquakes. As earthquakes occur in 
the future, additional cases of liquefaction-related damage must be expected. Design to prevent damage 
due to liquefaction consists of three parts: evaluation of liquefaction hazard, evaluation of potential 
ground displacement, and mitigating the hazard by designing to resist ground displacement or strength 
loss, by reducing the potential for liquefaction, or by choosing an alternative site with less hazard. 
Evaluation of liquefaction hazard. Liquefaction hazard at a site is commonly expressed in terms of a 
factor of safety. This factor is defined as the ratio between the available liquefaction resistance, 
expressed in terms of the cyclic stresses required to cause liquefaction, and the cyclic stresses generated 
by the design earthquake. Both of these stress parameters are commonly normalized with respect to the 
effective overburden stress at the depth in question to define a cyclic resistance ratio (CRR) and a cyclic 
stress ratio induced by the earthquake (CSR). 
The following possible methods for calculating the factor of safety against liquefaction have been 
proposed and used to various extents: 
1. Empirical MethodsThe most widely used method in practice involves empirical procedures. 
These procedures rely on correlations between observed cases of liquefaction and measurements 
made in the field with conventional exploration methods. Seed and Idriss (1971) first published the 
widely used simplified procedure utilizing the Standard Penetration Test (SPT). Since then, the 
procedure has evolved, primarily through summary papers by Professor H.B. Seed and his 
colleagues, and field test methods in addition to the SPT have been utilized in similar simplified 
procedures. These methods include cone penetrometer tests (CPTs), Becker hammer tests (BHTs), 
and shear wave velocity tests (SVTs). In 1996, a workshop sponsored by the National Center for 
Earthquake Engineering Research (NCEER) was convened by Professors T.L. Youd and I.M. Idriss 
with 20 experts to review and update the simplified procedure which had last been updated in 1985. 
The update of the simplified procedure that resulted from the NCEER workshop (termed herein the 
Liquefaction Workshop) is summarized in NCEER (1997) and in Youd et al. (2001). Martin and 
Lew (1999) focused on the implementation of this procedure in engineering practice, especially for 
southern California. The procedure described in NCEER (1997), Youd et al. (2001), and Martin 
and Lew (1999) using the Standard Penetration Test (SPT) is later summarized in this section. 
2. Analytical MethodsAnalytical methods are used less frequently to evaluate liquefaction potential 
 though they may be required for special projects or where soil conditions are not amenable to the 
empirical method. Analytical methods will also likely gain prominence with time as numerical 
methods and soil models improve and are increasingly validated. Originally (circa 1970s) the 
analytical method involved determination of the induced shearing stresses with a program such as 
SHAKE and comparing these stresses to results of cyclic triaxial or cyclic simple shear tests. Now 
the analytical method usually refers to a computer code that incorporates a soil model that calculates 
the buildup in pore water pressure. These more rigorous numerical methods include onedimensional, 
nonlinear effective stress codes such as DESRA and SUMDES and two dimensional, 
nonlinear effective stress codes such as FLAC, TARA, and DYNAFLOW. This new generation of 
analytical methods has soil models that are fit to laboratory data or liquefaction curves derived from 
SPT information. The methods are limited by the ability to represent the soil model from either the 
laboratory or field measurements and by the complexity of the wave propagation mechanisms, 
including the ability to select appropriate earthquake records to use in the analyses. 
3. Physical ModelingThese methods typically involved the use of centrifuges or relatively smallscale 
shaking tables to simulate seismic loading under well defined boundary conditions. More 
recently these methods have been expanded to include large laminar boxes mounted on very large 
Foundation Design Requirements 
143 
shake tables and full-scale field blast loading tests. Physical modeling of liquefaction is one of the 
main focus areas of the 2004-2014 Network for Earthquake Engineering Simulation (NEES) 
supported by the National Science Foundation. Soil used in the small-scale and laminar box models 
is reconstituted to represent different density and geometrical conditions. Because of difficulties in 
precisely modeling in-situ conditions at liquefiable sites, small-scale and laminar box models have 
seldom been used in design studies for specific sites. However, physical models are valuable for 
analyzing and understanding generalized soil behavior and for evaluating the validity of constitutive 
models under well defined boundary conditions. Recently, blast loading tests have been conducted 
to capture the in situ characteristics of the soil for research purposes (e.g., Treasure Island, 
California and in Japan). However, the cost and safety issues of this approach limits its use to only 
special design or research projects. 
The empirical approach for evaluating liquefaction hazards based on the Liquefaction Workshop and 
described in NCEER (1997) and Youd it al. (2001) is summarized in the following paragraphs. 
The first step in the liquefaction hazard evaluation using the empirical approach is usually to define the 
normalized cyclic shear stress ratio (CSR) from the peak horizontal ground acceleration expected at the 
site. This evaluation is made using the following simple equation: 
CSR = 0.65 (a max g )(sos 'o)r d (C7.4-1) 
where (amax/g) = peak horizontal acceleration at ground surface expressed as a decimal fraction of 
gravity, s0 = the vertical total stress in the soil at the depth in question, s0' = the vertical effective stress 
at the same depth, and rd = deformation-related stress reduction factor. 
The peak ground acceleration, amax, commonly used in liquefaction analysis is that which would occur at 
the site in the absence of liquefaction. Thus, the amax used in Eq. C7.4-1 is the estimated rock 
acceleration corrected for soil site response but with neglect of excess pore-water pressures that might 
develop. The acceleration can be determined using the general procedure described in Sec. 3.3 and 
taking amax equal to SDS/2.5. Alternatively, amax can be estimated from: (1) values obtained from the 
USGS national ground motion maps [see internet website http://geohazards.cr.usgs.gov/eq/] for a 
selected probability of exceedance, with correction for site effects using the Fa site factor in Sec. 3.3; or 
(2) from a site-specific ground motion analysis conforming to the requirements of Sec. 3.4. 
The stress reduction factor, rd, used in Eq. C7.4-1 was originally determined using a plot developed by 
Seed and Idriss (1971) showing the reduction factor versus depth. The consensus from the Liquefaction 
Workshop was to represent rd by the following equations: 
rd = 1.0  0.00765z for z = 9.15 m 
(C7.4-2) 
rd = 1.174  0.267z for 9.15 m < z = 23 m 
It should be noted that because nearly all the field data used to develop the simplified procedure are for 
depths less than 12 m, there is greater uncertainty in the evaluations at greater depths. The second step 
in the liquefaction hazard evaluation using the empirical approach usually involves determination of the 
normalized cyclic resistance ratio (CRR). The most commonly used empirical relationship for 
determining CRR was originally compiled by Seed et al. (1985). This relationship compares CRR with 
corrected Standard Penetration Test (SPT) resistance, (N1)60, from sites where liquefaction did or did not 
develop during past earthquakes. Figure C7.4-3 shows this relationship for Magnitude 7.5 earthquakes, 
with an adjustment at low values of CRR recommended by the Liquefaction Workshop. Similar 
relationships have been developed for determining CRR from CPT soundings, from BHT blowcounts, 
and from shear wave velocity data, as discussed by Youd et al. (2001) and as presented in detail in 
NCEER (1997). Only the SPT method is presented herein because of its more common use. 

2003 Commentary, Chapter 7 
144 
Figure C7.4-3. SPT clean sand base curve for magnitude 7.5 earthquakes with data from 
liquefaction case histories. (Modified from Seed et al., 1985). (NCEER, 1997; Youd et al., 2001). 
In Figure C7.4-3, CRRs calculated for various sites are plotted against (N1)60, where (N1)60 is the SPT 
blowcount normalized for an overburden stress of 100 kPa and for an energy ratio of 60 percent. Solid 
symbols represent sites where liquefaction occurred and open symbols represent sites where surface 
evidence of liquefaction was not found. Curves were drawn through the data to separate regions where 
liquefaction did and did not develop. As shown, curves are given for soils with fines contents (FC) 
ranging from less than 5 to 35 percent. 
While Figure C7.4-3 provides information about the variation in CRR with fines content, the preferred 
approach from the Liquefaction Workshop for adjusting for fines is to correct (N1)60 to an equivalent 
clean sand value, (N1)60cs using the following equations: 
(N1)60cs = a + (N1)60 (C7.4-3) 
where a and  = coefficients determined from the following relationships: 
a = 0 for FC = 5% 
a = exp[1.76  (190/FC2)] for 5% < FC < 35% 
a = 5.0 for FC = 35% 
Foundation Design Requirements 
145 
 = 1.0 for FC = 5% 
 = [0.99 + (FC1.5/1,000) for 5% < FC < 35% 
 = 1.2 for FC = 35% 
Several other corrections are made to (N1)60, as represented in the following equation: 
(N1)60 = NmCNCECBCRCS (C7.4-4) 
where Nm = measured standard penetration resistance; CN = factor to normalize Nm to a common 
reference effective overburden stress; CE = correction for hammer energy ratio (ER); CB = correction 
factor for borehole diameter; CR = correction factor for rod length; and CS = correction for samples with 
or without liners. Values given in Youd, et al., 2001) are shown in Table C.4-1. An alternative 
equation for Cn from that shown in the table (Youd, et al., 2001): 
CN = 2.2 / [1.2 + s'vo/Pa] (C7.4-5) 
where the maximum value of CN is equal to 1.7. The effective vertical stress, s'vo, is the stress at the 
time of the SPT measurement. Youd et al. (2001) caution that other means should be used to evaluate 
CN if s'vo is greater than 300 kPa. 
TABLE C7.4-1 
Corrections to SPT modified from Skempton 1986) as listed by Robertson and Wride (1998) (from 
Youd et al., 2001) 
Factor Equipment variable Term Correction 
Overburden pressure --- CN (Pa/s'vo) 0.5 
Overburden pressure --- CN CN < 1.7 
Energy ratio Donut hammer CE 0.5  1.0 
Energy ratio Safety hammer CE 0.7  1.2 
Energy ratio Automatic-trip Donuttype 
hammer 
CE 0.8  1.3 
Borehole diameter 65  115 mm CB 1.0 
Borehole diameter 150 mm CB 1.05 
Borehole diameter 200 mm CB 1.15 
Rod length < 3 CR 0.75 
Rod length 3  4 m CR 0.8 
Rod length 4 -6 m CR 0.85 
Rod length 6  10 m CR 0.95 
Rod length 10  30 m CR 1.0 
Sampling method Standard sampler CS 1.0 
Sampling method Sampler without liners CS 1.1  1.3 
It is very important that the engineer consider these correction factors when conducting the liquefaction 
analyses. Failure to consider these corrections can result in inaccurate liquefaction estimates  leading 
2003 Commentary, Chapter 7 
146 
to either excessive cost to mitigate the liquefaction concern or excessive risk of poor performance 
during a design event  potentially resulting in unacceptable damage. 
Special mention also needs to be made of the energy calibration term, CE. This correction has a very 
significant effect on the (N1)60 used to compute CRR. The value of this correction factor can vary 
greatly depending on the SPT hammer system used in the field and on site conditions. For important 
sites where CE could result in changes from liquefied to non liquefied, energy ratio measurements 
should be made. These measurements are relatively inexpensive and represent a small increase in 
overall field exploration costs. Many drilling contractors in areas that are seismically active provide 
calibrated equipment as part of their routine service. 
Before computing the factor of safety from liquefaction, the CRR result obtained from Figure C7.4-3 
(using the corrected SPT blow count identified in the equation for (N1)60 ) must be corrected for 
earthquake magnitude M if the magnitude differs from 7.5. The magnitude correction factor is shown in 
Figure C7.4-4. This plot was developed during the Liquefaction Workshop on the basis of input from 
experts attending the workshop. The range shown in Figure C7.4-4 is used because of uncertainties. 
The user should select a value consistent with the project risk. For M greater than 7.5 the factors 
recommended by Idriss (second from highest) should be used. 
Figure C7.4-4. Magnitude scaling factors derived by various investigators. (NCEER, 1997; Youd 
et al., 2001) 
The magnitude, M, needed to determine a magnitude scaling factor from Figure C7.4-4 should 
correspond to the Maximum Considered Earthquake (MCE). Where the general procedure for ground 
motion estimation is used (Sec. 3.3) and the MCE is determined probabilistically, the magnitude used in 
these evaluations can be obtained from deaggregation information available by latitude and longitude 
from the USGS website (http://geohazards.cr.usgs.gov/eq/). Where the general procedure (Sec. 3.3) is 
used and the MCE is bounded deterministically near known active fault sources (Commentary Appendix 
A), the magnitude of the MCE should be the characteristic maximum magnitude assigned to the fault in 
the construction of the MCE ground motion maps. Where the site-specific procedure for ground motion 
estimation is used (Sec. 3.4), the magnitude of the MCE should be similarly determined from the sitespecific 
analysis. In all cases, it should be remembered that the likelihood of liquefaction at the site (as 
defined later by the factor of safety FL in Eq. C7.4-6) is determined jointly by amax and M and not by 
amax alone. Because of the longer duration of strong ground-shaking, large distant earthquakes may in 
Foundation Design Requirements 
147 
some cases generate liquefaction at a site while smaller nearby earthquakes may not generate 
liquefaction even though amax of the nearer events is larger than that from the more distant events. 
The final step in the liquefaction hazard evaluation using the empirical approach involves the 
computation of the factor of safety (FL) against liquefaction using the equation: 
FL = CRR/CSR (C7.4-6) 
If FL is greater than one, then liquefaction should not develop. If at any depth in the sediment profile, 
FL is equal to or less than one, then there is a liquefaction hazard. Although the curves shown in Figure 
C7.4-3 envelop the plotted data, it is possible that liquefaction may have occurred beyond the enveloped 
data and was not detected at ground surface. For this reason a factor of safety of 1.2 to 1.5 is usually 
appropriate for building sites  with the actual factor selected on the basis of the importance of the 
structure and the potential for ground displacement at the site. 
Additional guidance on the selection of the appropriate factor of safety is provided by Martin and Lew 
1999. They suggest that the following factors be considered when selecting the factor of safety: 
1. The type of structure and its vulnerability to damage. 
2. Levels of risk accepted by the owner or governmental regulations with questions related to design 
for life safety, limited structural damage, or essentially no damage. 
3. Damage potential associated with the particular liquefaction hazards. Flow failures or major lateral 
spreads pose more damage potential than differential settlement. Hence factors of safety could be 
adjusted accordingly. 
4. Damage potential associated with design earthquake magnitude. A magnitude 7.5 event is 
potentially more damaging than a 6.5 event. 
5. Damage potential associated with SPT values; low blow counts have a greater cyclic strain 
potential than higher blowcounts. 
6. Uncertainty in SPT- or CPT- derived liquefaction strengths used for evaluations. Note that a change 
in silt content from 5 to 15 percent could change a factor of safety from, say, 1.0 to 1.25. 
7. For high levels of design ground motion, factors of safety may be indeterminate. For example, if 
(N1)60) = 20, M = 7.5, and fines content = 35 percent, liquefaction strengths cannot be accurately 
defined due to the vertical asymptote on the empirical strength curve. 
Martin and Lew (1999) indicate that the final choice of an appropriate factor of safety must reflect the 
particular conditions associated with the specific site and the vulnerability of site-related structures. 
Table C7.4-2 summarizes factors of safety suggested by Martin and Lew. 
Table C7.4-2. Factors of safety for liquefaction hazard assessment (from Martin and Lew, 
1999). 
Consequences of 
Liquefaction 
(N1)60)cs Factor of Safety 
Settlement = 15 1.1 
= 30 1.0 
Surface Manifestations = 15 1.2 
= 30 1.0 
Lateral Spread = 15 1.3 
= 30 1.0 
2003 Commentary, Chapter 7 
148 
As a final comment on the assessment of liquefaction hazards, it is important to note that soils 
composed of sands, silts, and gravels are most susceptible to liquefaction while clayey soils generally 
are not susceptible to this phenomenon. The curves in Figure C7.4-3 are valid for soils composed 
primarily of sand. The curves should be used with caution for soils with substantial amounts of gravel. 
Verified corrections for gravel content have not been developed; a geotechnical engineer, experienced in 
liquefaction hazard evaluation, should be consulted when gravelly soils are encountered. For soils 
containing more than 35 percent fines, the curve in Figure C7.4-3 for 35 percent fines should be used 
provided the following criteria are met (Seed and Idriss, 1982; Seed et al., 1983): the weight of soil 
particles finer than 0.005 mm is less than 15 percent of the dry weight of a specimen of the soil; the 
liquid limit of soil is less than 35 percent; and the moisture content of the in-place soil is greater than 0.9 
times the liquid limit. If these criteria are not met, the soils may be considered nonliquefiable. 
Evaluation of potential for ground displacements. Liquefaction by itself may or may not be of 
engineering significance. Only when liquefaction is accompanied by loss of ground support and/or 
ground deformation does this phenomenon become important to structural design. Surface 
manifestations, loss of bearing strength, ground settlement, flow failure and lateral spread are ground 
failure mechanisms that have caused structural damage during past earthquakes. These types of ground 
failure are described in Martin and Lew (1999), U.S. Army Corps of Engineers (1998) and National 
Research Council (1985) and are discussed below. The type of failure and amount of ground 
displacement are a function of several parameters including the looseness of the liquefied soil layer, the 
thickness and extent of the liquefied layer, the thickness and permeability of unliquefied material 
overlying the liquefied layer, the ground slope, and the nearness of a free face. 
Surface Manifestations. Surface manifestations refer to sand boils and ground fissures on level ground 
sites. For structures supported on shallow foundations, the effects of surface manifestations on the 
structure could be tilting or cracking. Criteria are given by Ishihara (1985) for evaluating the influence 
of thickness of layers on surface manifestation of liquefaction effects for level sites. These criteria may 
be used for noncritical or nonessential structures on level sites not subject to lateral spreads (see later in 
this section). Additional analysis should be performed for critical or essential structures. 
Loss of bearing strength. Loss of bearing strength can occur if the foundation is located within or above 
the liqefiable layer. The consequence of bearing failure could be settlement or tilting of the structure. 
Usually, loss of bearing strength is not likely for light structures with shallow footings founded on 
stable, nonliquefiable materials overlying deeply buried liquefiable layers, particularly if the liquefiable 
layers are relatively thin. Simple guidance for how deep or how thin the layers must be has not yet been 
developed. Martin and Lew (1999) provide some preliminary guidance based on the Ishihara (1985) 
method. Final evaluation of the potential for loss of bearing strength should be made by a geotechnical 
engineer experienced in liquefaction hazard assessment 
Ground settlement. For saturated or dry granular soils in a loose condition, the amount of ground 
settlement could approach 3 to 4 percent of the thickness of the loose soil layer in some cases. This 
amount of settlement could cause tilting or cracking of a building, and therefore, it is usually important 
to evaluate the potential for ground settlement during earthquakes. 
Tokimatsu and Seed (1987) published an empirical procedure for estimating ground settlement. It is 
beyond the scope of this commentary to outline that procedure which, although explicit, has several 
rather complex steps. The Tokimatsu and Seed procedure can be applied whether liquefaction does or 
does not occur. For dry cohesionless soils, the settlement estimate from Tokimatsu and Seed should be 
multiplied by a factor of 2 to account for multi-directional shaking effects as discussed by Martin and 
Lew (1999). 
Flow failures. Flow failures or flow slides are the most catastrophic form of ground failure that may be 
triggered when liquefaction occurs. They may displace large masses of soils tens of meters. Flow slides 
occur when the average static shear stresses on potential failure surfaces are less than the average shear 
strengths of liquefied soil on these surfaces. Standard limit equilibrium static slope stability analyses 
may be used to assess flow failure potential with the residual strength of liquefied soil used as the 
strength parameter in the analyses. 
Foundation Design Requirements 
149 
The determination of residual strengths is very inexact, and consensus as to the most appropriate 
approach has not been reached to date. Two relationships for residual strength of liquefied soil that are 
often used in practice are those of Seed and Harder (1990) and Stark and Mesri (1992). A more complete 
discussion and references on this topic may be found in Martin and Lew (1999). 
Lateral spreads. Lateral spreads are ground-failure phenomena that can occur on gently sloping ground 
underlain by liquefied soil. They may result in lateral movements in the range of a few centimeters to 
several meters. Earthquake ground-shaking affects the stability of gently sloping ground containing 
liquefiable materials by seismic inertia forces combined with static gravity forces within the slope and 
by shaking-induced strength reductions in the liquefiable materials. Temporary instability due to 
seismic inertia forces are manifested by lateral downslope movement. For the duration of ground 
shaking associated with moderate-to large-magnitude earthquakes, there could be many such 
occurrences of temporary instability during earthquake shaking, producing an accumulation of 
downslope movement. 
Various analytical and empirical techniques have been developed to date to estimate lateral spread 
ground displacement; however, no single technique has been widely accepted or verified for 
engineering design. Three approaches are used depending on the requirements of the project: empirical 
procedures, simplified analytical methods, and more rigorous computer modeling. Empirical 
procedures use correlations between past ground displacement and site conditions under which those 
displacements occurred. Youd et al. (2002) present an empirical method that provides an estimate of 
lateral spread displacements as a function of earthquake magnitude, distance, topographic conditions, 
and soil deposit characteristics. As shown in Figure C7.4-5, the displacements estimated by the Youd et 
al. (2002) method are generally within a factor of two of the observed displacements. Bardet et al. 
(2002) present an empirical method having a formulation similar to that of Youd et al. (2002) but using 
fewer parameters to describe the soil deposit. The Bardet et al. (2002) model was developed to assess 
lateral spread displacements at a regional scale rather than for site-specific applications. 
Simplified analytical techniques generally apply some form of Newmarks analysis of a rigid body 
sliding on an infinite or circular failure surface with ultimate shear resistance estimated from the 
residual strength of the deforming soil. Additional discussion of the simplified Newmark method is 
provided in Sec. 7.4.1. More rigorous computer modeling typically involves use of nonlinear finite 
element or finite difference methods to predict deformations, such as with the computer code FLAC. 
Both the simplified Newmark method and the rigorous computer codes require considerable experience 
to obtain meaningful results. For example, the soil model within the nonlinear computer codes is often 
calibrated for only specific conditions. If the site is not characterized by these conditions, errors in 
estimating the displacement by a factor of two or more can easily occur. 

2003 Commentary, Chapter 7 
150 
Figure C7.4-5. Measured versus predicted displacements for displacements up to 2 meters. (Youd 
et al., 2002). 
Liquefaction-induced deformations are not directly proportional to ground motions and may be more 
than 50 percent higher for maximum considered earthquake ground motions than for design earthquake 
ground motions. The liquefaction potential and resulting deformations for ground motions consistent 
with the maximum considered earthquake should also be evaluated and, while not required in the 
Provisions, should be used by the registered design professional in checking for building damage that 
may result in collapse. In addition, Seismic Use Group III structures should be checked for their 
required post-earthquake condition. 
Mitigation of liquefaction hazard. With respect to the hazard of liquefaction, three mitigative 
measures might be considered: design the structure to resist the hazard, stabilize the site to reduce the 
hazard, or choose an alternative site. Structural measures that are used to reduce the hazard include 
deep foundations, mat foundations, or footings interconnected with ties as discussed in Sec. 7.4.3. Deep 
foundations have performed well at level sites of liquefaction where effects were limited to ground 
settlement and ground oscillation with no more than a few inches of lateral displacement. Deep 
foundations, such as piles, may receive reduced soil support through the liquefied layer and may be 
subjected to transient lateral displacements across the layer. Well reinforced mat foundations also have 
performed well at localities where ground displacements were less than 1 ft, although re-leveling of the 
structure has been required in some instances (Youd, 1989). Strong ties between footings also should 
provide increased resistance to damage where differential ground displacements are less than a foot. 
Evaluations of structural performance following two recent Japanese earthquakes, 1993 Hokkaido 
Nansei-Oki and 1995 (Kobe) Hyogo-Ken Nanbu, indicate that small structures on shallow foundations 
performed well in liquefaction areas. Sand boil eruptions and open ground fissures in these areas 
indicate minor effects of liquefaction, including ground oscillation and up to several tenths of a meter of 
lateral spread displacement. Many small structures (mostly houses, shops, schools, etc.) were 
Foundation Design Requirements 
151 
structurally undamaged although a few tilted slightly. Foundations for these structures consist of 
reinforced concrete perimeter wall footings with reinforced concrete interior wall footings tied into the 
perimeter walls at intersections. These foundations acted as diaphragms causing the soil to yield 
beneath the foundation which prevented fracture of foundations and propagation of differential 
displacements into the superstructure. 
Similarly, well reinforced foundations that would not fracture could be used in U.S. practice as a 
mitigative measure to reduce structural damage in areas subject to liquefaction but with limited potential 
for lateral (< 0.3 m) or vertical (< 0.05 m) ground displacements. Such strengthening also would serve 
as an effective mitigation measure against damage from other sources of limited ground displacement 
including fault zones, landslides, and cut fill boundaries. Where slab-on-grade or basement slabs are 
used as foundation elements, these slabs should be reinforced and tied to the foundation walls to give 
the structure adequate strength to resist ground displacement. Although strengthening of foundations, as 
noted above, would largely mitigate damage to the structure, utility connections may be adversely 
affected unless special flexibility is built into these nonstructural components. 
Another possible consequence of liquefaction to structures is increased lateral pressures against 
basement walls as discussed in Sec. 7.5.1. A common procedure used in design for such increased 
pressures is to assume that the liquefied material acts as a dense fluid having a unit weight of the 
liquefied soil. The wall then is designed assuming that hydrostatic pressure for the dense fluid acts 
along the total subsurface height of the wall. The procedure applies equivalent horizontal earth 
pressures that are greater than typical at-rest earth pressures but less than passive earth pressures. As a 
final consideration, to prevent buoyant rise as a consequence of liquefaction, the total weight of the 
structure should be greater than the volume of the basement or other cavity times the unit weight of 
liquefied soil. (Note that structures with insufficient weight to counterbalance buoyant effects could 
differentially rise during an earthquake.) 
At sites where expected ground displacements are unacceptably large, ground modification to lessen the 
liquefaction or ground failure hazard or selection of an alternative site may be required. Techniques for 
ground stabilization to prevent liquefaction of potentially unstable soils include removal and 
replacement of soil; compaction of soil in place using vibrations, heavy tamping, compaction piles, or 
compaction grouting; buttressing; chemical stabilization with grout; and installation of drains. Further 
explanation of these methods is given by the National Research Council (1985). 
Surface fault rupture hazard. Fault ruptures during past earthquakes have led to large surface 
displacements that are potentially destructive to engineered construction. Displacements, which range 
from a fraction of an inch to tens of feet, generally occur along traces of previously active faults. The 
sense of displacement ranges from horizontal strike-slip to vertical dip-slip to many combinations of 
these components. The following commentary summarizes procedures to follow or consider when 
assessing the hazard of surface fault rupture. Sources of detailed information for evaluating the hazard 
of surface fault rupture include Slemmons and dePolo (1986), the Utah Section of the Association of 
Engineering Geologists (1987), Swan et al. (1991), Hart and Bryant (1997), and California Geological 
Survey (2002). Other beneficial references are given in the bibliographies of these publications. 
Assessment of surface faulting hazard. The evaluation of surface fault rupture hazard at a given site is 
based extensively on the concepts of recency and recurrence of faulting along existing faults. The 
magnitude, sense, and frequency of fault rupture vary for different faults or even along different 
segments of the same fault. Even so, future faulting generally is expected to recur along pre-existing 
active faults. The development of a new fault or reactivation of a long inactive fault is relatively 
uncommon and generally need not be a concern. For most engineering applications related to 
foundation design, a sufficient definition of an active fault is given in CDMG Special Publication 42 
(Hart and Bryant, 1997): An active fault has had displacement in Holocene time (last 11,000 years). 
As a practical matter, fault investigations should be conducted by qualified geologists and directed at the 
problem of locating faults and evaluating recency of activity, fault length, the amount and character of 
past displacements, and the expected amount and potential of future displacement. Identification and 
2003 Commentary, Chapter 7 
152 
characterization studies should incorporate evaluation of regional fault patterns as well as detailed study 
of fault features at and in the near vicinity (within a few hundred yards to a mile) of the site. Detailed 
studies often include trenching to accurately locate, document, and date fault features. 
Suggested approach for assessing surface faulting hazard. The following approach should be used, or 
at least considered, in fault hazard assessment. Some of the investigative methods outlined below 
should be carried out beyond the site being investigated. However, it is not expected that all of the 
following methods would be used in a single investigation: 
1. A review should be made of the published and unpublished geologic literature from the region 
along with records concerning geologic units, faults, ground-water barriers, etc. 
2. A stereoscopic study of aerial photographs and other remotely sensed images should be made to 
detect fault-related topography/geomorphic features, vegetation and soil contrasts, and other 
lineaments of possible fault origin. The study of predevelopment aerial photographs is often 
essential to the detection of fault features. 
3. A field reconnaissance study generally is required and should include observation and mapping of 
bedrock and soil units and structures, geomorphic surfaces, fault-related geomorphic features, 
springs, and deformation of man-made structures due to fault creep. Field study should be detailed 
within the site with less detailed reconnaissance of an area within a mile or so of the site. 
4. Subsurface investigations usually are needed to evaluate location and activity of fault traces. These 
investigations may include trenches, test pits, and/or boreholes to permit detailed and direct 
observation of geologic units and faults. 
5. The geometry of faults may be further defined by geophysical investigations including seismic 
refraction, seismic reflection, gravity, magnetic intensity, resistivity, ground penetrating radar, etc. 
These indirect methods require a knowledge of specific geologic conditions for reliable 
interpretation. Geophysical methods alone never prove the absence of a fault and they typically do 
not identify the recency of activity. 
6. More sophisticated and more costly studies may provide valuable data where geological special 
conditions exist or where requirements for critical structures demand a more intensive investigation. 
These methods might involve repeated geodetic surveys, strain measurements, or monitoring of 
microseismicity and radiometric analysis (C14, K-Ar), stratigraphic correlation (fossils, 
mineralology) soil profile development, paleomagnetism (magnetostratigraphy), or other dating 
techniques (thermoluminescense, cosmogenic isotopes) to date the age of faulted or unfaulted units 
or surfaces. Probabilistic studies may be considered to evaluate the probability of fault 
displacement (Youngs et al., 2003). 
The following information should be developed to provide documented support for conclusions relative 
to location and magnitude of faulting hazards: 
1. Maps should be prepared showing the existence (or absence) and location of hazardous faults on or 
near the site. The distribution of primary and secondary faulting (fault zone width) and fault-related 
surface deformation should be shown. 
2. The type, amount, and sense of displacement of past surface faulting episodes should be 
documented, if possible. 
3. From this documentation, estimates of location, magnitude, and likelihood or relative potential for 
future fault displacement can be made, preferably from measurements of past surface faulting events 
at the site, using the premise that the general pattern of past activity will repeat in the future. 
Estimates also may be made from empirical correlations between fault displacement and fault length 
or earthquake magnitude published by Wells and Coppersmith (1994). Where fault segment length 
and sense of displacement are defined, these correlations may provide an estimate of future fault 
displacement (either the maximum or the average to be expected). 
4. The degree of confidence and limitations of the data should be addressed. 
Foundation Design Requirements 
153 
There are no codified procedures for estimating the amount or probability of future fault displacements. 
Estimates may be made, however, by qualified earth scientists using techniques described above. 
Because techniques for making these estimates are not standardized, peer review of reports is useful to 
verify the adequacy of the methods used and the estimates reports, to aid the evaluation by the 
permitting agency, and to facilitate discussion between specialists that could lead to the development of 
standards. 
The following guidelines are given for safe siting of engineered construction in areas crossed by active 
faults: 
1. Where ordinances have been developed that specify safe setback distances from traces of active 
faults or active fault zones, those distances must be complied with and accepted as the minimum for 
safe siting of buildings. For example, the general setback requirement in California is a minimum 
of 50 ft from a well-defined zone containing the traces of an active fault. That setback distance is 
mandated as a minimum for structures near faults unless a site-specific special geologic 
investigation shows that a lesser distance could be safety applied (California Code of Regulations, 
Title 14, Division 2, Sec. 3603(a)). 
2. In general, safe setback distances may be determined from geologic studies and analyses as noted 
above. Setback requirements for a site should be developed by the site engineers and geologists in 
consultation with professionals from the building and planning departments of the jurisdiction 
involved. Where sufficient geologic data have been developed to accurately locate the zone 
containing active fault traces and the zone is not complex, a 50-ft setback distance may be specified. 
For complex fault zones, greater setback distances may be required. Dip-slip faults, with either 
normal or reverse motion, typically produce multiple fractures within rather wide and irregular fault 
zones. These zones generally are confined to the hanging-wall side of the fault leaving the footwall 
side little disturbed. Setback requirements for such faults may be rather narrow on the footwall 
side, depending on the quality of the data available, and larger on the hanging wall side of the zone. 
Some fault zones may contain broad deformational features such as pressure ridges and sags rather 
than clearly defined fault scarps or shear zones. Nonessential structures may be sited in these zones 
provided structural mitigative measures are applied as noted below. Studies by qualified geologists 
and engineers are required for such zones to assure that building foundations can withstand probable 
ground deformations in such zones. 
Mitigation of surface faulting hazards. There is no mitigative technology that can be used to prevent 
fault rupture from occurring. Thus, sites with unacceptable faulting hazard must either be avoided or 
structures designed to withstand ground deformation or surface fault rupture. In general practice, it is 
economically impractical to design a structure to withstand more than a few inches of fault 
displacement. Some buildings with strong foundations, however, have successfully withstood or 
diverted a few inches of surface fault rupture without damage to the structure (Youd, 1989; Kelson et 
al., 2001). Well reinforced mat foundations and strongly inter-tied footings have been most effective. 
In general, less damage has been inflicted by compressional or shear displacement than by vertical or 
extensional displacements. 
7.4.2 Pole-type structures. The use of pole-type structures is permitted. These structures are 
inherently sensitive to earthquake motions. Bending in the poles and the soil capacity for lateral 
resistance of the portion of the pole embedded in the ground should be considered and the design 
completed accordingly. 
7.4.3 Foundation ties. One of the prerequisites of adequate performance of a building during an 
earthquake is the provision of a foundation that acts as a unit and does not permit one column or wall to 
move appreciably with respect to another. A common method used to attain this is to provide ties 
between footings and pile caps. This is especially necessary where the surface soils are soft enough to 
require the use of piles or caissons. Therefore, the pile caps or caissons are tied together with nominal 
ties capable of carrying, in tension or compression, a force equal to SDS/10 times the larger pile cap or 
column load. 
2003 Commentary, Chapter 7 
154 
A common practice in some multistory buildings is to have major columns that run the full height of the 
building adjacent to smaller columns in the basement that support only the first floor slab. The 
coefficient applies to the heaviest column load. 
Alternate methods of tying foundations together are permitted (such as using a properly reinforced floor 
slab that can take both tension and compression). Lateral soil pressure on pile caps is not a 
recommended method because the motion is imparted from soil to structure (not inversely as is 
commonly assumed), and if the soil is soft enough to require piles, little reliance can be placed on 
soft-soil passive pressure to restrain relative displacement under dynamic conditions. 
If piles are to support structures in the air or over water (such as in a wharf or pier), batter piles may be 
required to provide stability or the piles may be required to provide bending capacity for lateral 
stability. It is up to the foundation engineer to determine the fluidity or viscosity of the soil and the 
point where lateral buckling support to the pile can be provided (that is, the point where the flow of the 
soil around the piles may be negligible). 
7.4.4 Special pile requirements. Special requirements for piles, piers, or caissons in Seismic Design 
Category C are given in this section. Provisions for pile anchorage to the pile cap or grade beam and 
transverse reinforcement detailing requirements for concrete piles are provided. The anchorage 
requirements are intended to assure that the connection to the pile cap does not fail in a brittle manner 
under moderate ground motions. Moderate ground motions could result in pile tension forces or 
bending moments which could compromise shallow anchorage embedment. Shallow anchorages in pile 
caps may consist of short lengths of reinforcing bars or bare structural steel pile sections. Loss of pile 
anchorage could result in unintended increases in vertical seismic force resisting element drifts from 
rocking, potential overturning instability of the superstructure, and loss of shearing resistance at the 
ground surface. Anchorage by shallow embedment of the bare steel pile section is not recommended 
due to the degradation of the concrete bond from cracking as a result of the cyclic loading from the 
moderate ground motions. Exception to this is permitted for steel pipe piles filled with concrete when 
the connection is made with reinforcing bar dowels properly developed into the pile and pile cap. The 
confinement of the interior concrete by the hoop stresses of the circular pile section was judged to be 
sufficient to prevent concrete pullout from that section. Using this method of connection, the structural 
steel pipe section should be embedded into the pile cap for a short distance or else the pile should be 
designed as an uncased concrete pile. End anchorage detailing requirements for transverse 
reinforcement generally follow that required by ACI 318, Chapter 21 to assure that no loss of 
confinement of the transverse reinforcement occurs in concrete piles since verification of pile damage 
after moderate ground motions is difficult or not done. 
7.4.4.1 Uncased concrete piles. The uncased concrete piles category has been expanded to include 
auger-cast piles which are now subject to the same reinforcement requirements as other cast-in-place 
concrete piles. The longitudinal reinforcement within a pile has prescriptive termination point 
minimums intended for firm soils and is required to extend at least the full flexural length. The 
longitudinal reinforcement should extend past the flexural length by its development length 
requirement. The flexural length has been defined as that length of a pile from bottom of pile cap to a 
point where 0.4 times the concrete section cracking moment exceeds the calculated flexural demand. 
The 0.4 factor is analogous to a material resistance factor in strength design. This definition implies the 
plain concrete section will be resisting some minimal amount of moment demand beyond the flexural 
length. Where the pile is subject to significant uplift forces, it is recommended that the longitudinal 
reinforcement extend the full length of the pile. 
Increased transverse reinforcement requirements are given for the potential plastic hinge zone 
immediately below the pile cap. The potential plastic hinge zone was taken to be three pile diameters 
below the pile cap to allow for varied soil site classes. The transverse reinforcement detailing for this 
zone is similar to that required for concrete intermediate moment frames at hinge regions and is 
expected to provide a displacement ductility of approximately 4. Beyond the potential plastic hinge 
region, the curvature ductility demand is not considered to exceed that provided by the nominal moment 
capacity of the section for non-earthquake loads. 
Foundation Design Requirements 
155 
7.4.4.4 Precast (non-prestressed) concrete piles. For precast concrete piles, the longitudinal 
reinforcement is specified to extend the full length of the pile so there is no need to determine the 
flexural length. Transverse reinforcement spacing within the potential plastic hinge zone is required for 
the length of three pile diameters at the bottom of pile cap. Particular attention should be taken where 
piles cannot be driven to or are overdriven beyond the anticipated end bearing point elevation. The 
transverse reinforcement size and spacing in this region is the same as the uncased concrete pile. 
Transverse reinforcement spacing outside the potential plastic hinge zone is specified to be no greater 
than 8 inches to conform with current building code minimums for this pile type. 
7.4.4.5 Precast-prestressed piles. The transverse reinforcement requirements are primarily taken from 
the PCI Committee Report (1993) on precast prestressed concrete piling for geographic regions subject 
to low to moderate ground motions. The amount of transverse reinforcement was relaxed for the pile 
region greater than 20 feet (6m) below the pile cap to one-half of that required above. It was judged that 
the reduced transverse reinforcement would be sufficient to resist the reduced curvature demands at that 
point. Particular attention should be taken where piles cannot be driven to or are overdriven beyond the 
anticipated end bearing point elevation so that the length of the confining transverse reinforcement is 
maintained. 
Equation (7.4-1), originally from ACI 318, has always been intended to be a lower bound spiral 
transverse reinforcement ratio for larger diameter columns. It is independent of the member section 
properties and can therefore be applied to large or small diameter piles. For cast-in-place piles and 
prestressed concrete piles, the resulting spiral reinforcement ratios from this formula are considered to 
be sufficient to provide moderate ductility capacities. 
High strength hard drawn wire with higher yield strengths is permitted to be used for transverse circular 
spiral reinforcement of precast prestressed concrete piles. Pile test specimens using this type of 
transverse reinforcement include the research done by Park and Hoat Joen (1990). High strength hard 
drawn wire has yield strengths between 150 and 200 ksi. fyh is conservatively limited to 85 ksi for this 
steel because hard drawn wire has limited ductility. 
7.5 SEISMIC DESIGN CATEGORIES D, E, AND F 
For Seismic Design Category D, E, or F construction, all the preceding provisions for Seismic Design 
Category C applies for the foundations, but the earthquake detailing is generally more severe and 
demanding. 
7.5.1 Investigation. In addition to the potential site hazards discussed in Provisions Sec. 7.4.1, 
consideration of lateral pressures on earth retaining structures shall be included in investigations for 
Seismic Design Categories D, E, and F. 
Earth retaining structures. Increased lateral pressures on retaining structures during earthquakes have 
long been recognized; however, design procedures have not been prescribed in U.S. model building 
codes. Waterfront structures often have performed poorly in major earthquake due to excess pore water 
pressure and liquefaction conditions developing in relatively loose, saturated granular soils. Damage 
reports for structures away from waterfronts are generally limited with only a few cases of stability 
failures or large permanent movements (Whitman, 1991). Due to the apparent conservatism or 
overstrength in static design of most walls, the complexity of nonlinear dynamic soil-structure 
interaction, and the poor understanding of the behavior of retaining structures with cohesive or dense 
granular soils, Whitman (1991) recommends that engineers must rely primarily on a sound 
understanding of fundamental principles and of general patterns of behavior. 
Seismic design analysis of retaining walls is discussed below for two categories of walls: yielding 
walls that can move sufficiently to develop minimum active earth pressures and nonyielding walls 
that do not satisfy this movement condition. The amount of movement to develop minimum active 
pressure is very small. A displacement at the top of the wall of 0.002 times the wall height is typically 
sufficient to develop the minimum active pressure state. Generally, free-standing gravity or cantilever 
2003 Commentary, Chapter 7 
156 
walls are considered to be yielding walls (except massive gravity walls founded on rock), whereas 
building basement walls restrained at the top and bottom are considered to be nonyielding. 
Yielding walls. At the 1970 Specialty Conference on Lateral Stresses in the Ground and Design of 
Earth Retaining Structures, Seed and Whitman (1970) made a significant contribution by reintroducing 
and reformulating the Monobe-Okabe (M-O) seismic coefficient analysis (Monobe and Matsuo, 1929; 
Okabe, 1926), the earliest method for assessing the dynamic lateral pressures on a retaining wall. The 
M-O method is based on the key assumption that the wall displaces or rotates outward sufficiently to 
produce the minimum active earth pressure state. The M-O formulation is expressed as: 
PAE=(1/ 2).H2(1-kv)KAE (C7.5-1) 
where: PAE is the total (static + dynamic) lateral thrust, . is unit weight of backfill soil, H is height of 
backfill behind the wall, kV is vertical ground acceleration divided by gravitational acceleration, and KAE 
is the static plus dynamic lateral earth pressure coefficient which is dependent on (in its most general 
form) angle of friction of backfill, angle of wall friction, slope of backfill surface, and slope of back face 
of wall, as well as horizontal and vertical ground acceleration. The formulation for KAE is given in 
textbooks on soil dynamics (Prakash, 1981; Das, 1983; Kramer, 1996) and discussed in detail by 
Ebeling and Morrison (1992). 
Seed and Whitman (1970), as a convenience in design analysis, proposed to evaluate the total lateral 
thrust, PAE, in terms of its static component (PA) and dynamic incremental component (.PAE): 
PAE =PA +.PAE (C7.5-2a) 
or 
KAE =KA+.KAE (C7.5-2b) 
or 
.PAE =(1/ 2).H2.KAE (C7.5-2c) 
Seed and Whitman (1970), based on a parametric sensitivity analysis, further proposed that for practical 
purposes: 
.KAE =(3/ 4)Kh (C7.5-3a) 
.PAE =(1/ 2).H2(3/ 4)kh =(3/8)kh.H2 (C7.5-3b) 
where kh is horizontal ground acceleration divided by gravitational acceleration. It is recommended that 
kh be taken equal to the site peak ground acceleration that is consistent with design earthquake ground 
motions as determined in Provisions Sec. 7.5.2 (that is, kh = SDS/2.5). Eq. C7.5-3a and C7.5-3b 
generally are referred to as the simplified M-O formulation. 
Since its introduction, there has been a consensus in geotechnical engineering practice that the 
simplified M-O formulation reasonably represents the dynamic (seismic) lateral earth pressure 
increment for yielding retaining walls. For the distribution of the dynamic thrust, .PAE, Seed and 
Whitman (1970) recommended that the resultant dynamic thrust act at 0.6H above the base of the wall 
(that is, inverted trapezoidal pressure distribution). 
Using the simplified M-O formulation, a yielding wall may be designed using either a limit-equilibrium 
force approach (conventional retaining wall design) or an approach that permits movement of the wall 
up to tolerable amounts. Richards and Elms (1979) introduced a method for seismic design analysis of 
yielding walls considering translational sliding as a failure mode and based on tolerable permanent 
displacements for the wall. There are a number of empirical formulations for estimating permanent 
displacements under a translation mode of failure; these have been reviewed by Whitman and Liao 
(1985). Nadim (1980) and Nadim and Whitman (1984) incorporated the failure mode of wall tilting as 
well as sliding by employing coupled equations of motion, which were further formulated by 
Foundation Design Requirements 
157 
Siddharthan et al. (1992) as a design method to predict the seismic performance of retaining walls 
taking into account both sliding and tilting. Alternatively, Prakash and others (1995) described design 
procedures and presented design charts for estimating both sliding and rocking displacements of rigid 
retaining walls. These design charts are the results of analyses for which the backfill and foundation 
soils were modeled as nonlinear viscoelastic materials. A simplified method that considers rocking of a 
wall on a rigid foundation about the toe was described by Steedman and Zeng (1996) and allows the 
determination of the threshold acceleration beyond which the wall will rotate. A simplified procedure 
for evaluating the critical threshold accelerations for sliding and tilting was described by Richards and 
others (1996). 
Application of methods for evaluating tilting of yielding walls has been limited to a few case studies and 
back-calculation of laboratory test results. Evaluation of wall tilting requires considerable engineering 
judgment. Because the tilting mode of failure can lead to instability of a yielding retaining wall, it is 
suggested that this mode of failure be avoided in the design of new walls by proportioning the walls to 
prevent rotation in order to displace only in the sliding mode. 
Nonyielding walls. Wood (1973) analyzed the response of a rigid nonyielding wall retaining a 
homogeneous linear elastic soil and connected to a rigid base. For such conditions, Wood established 
that the dynamic amplification was insignificant for relatively low-frequency ground motions (that 
is, motions at less than half of the natural frequency of the unconstrained backfill), which would include 
many or most earthquake problems. 
For uniform, constant kh applied throughout the elastic backfill, Wood (1973) developed the dynamic 
thrust, .PE, acting on smooth rigid nonyielding walls as: 
2 
.PE=Fkh.H (C7.5-4a) 
The value of F is approximately equal to unity (Whitman, 1991) leading to the following approximate 
formulation for a rigid nonyielding wall on a rigid base: 
2 
.PE=kh.H (C7.5-4b) 
As for yielding walls, the point of application of the dynamic thrust is taken typically at a height of 0.6H 
above the base of the wall. 
It should be noted that the model used by Wood (1973) does not incorporate any effect on the pressures 
of the inertial response of a superstructure connected to the top of the wall. This effect may modify the 
interaction between the soil and the wall and thus modify the pressures from those calculated assuming 
a rigid wall on a rigid base. The subject of soil-wall interaction is addressed in the following sections. 
This section also provides further discussion on the applicability of the Wood and the M-O 
formulations. 
Soil-structure-interaction approach and modeling for wall pressures. Lam and Martin (1986), 
Soydemir and Celebi (1992), Veletsos and Younan (1994a and 1994b), and Ostadan and White (1998), 
among others, argue that the earth pressures acting on the walls of embedded structures during 
earthquakes are primarily governed by soil-structure interaction (SSI) and, thus, should be treated 
differently from the concept of limiting equilibrium (that is, M-O method). Soil-structure interaction 
includes both a kinematic componentthe interaction of a massless rigid wall with the adjacent soil as 
modeled by Wood (1973)and an inertial componentthe interaction of the wall, connected to a 
responding superstructure, with the adjacent soil. Detailed SSI analyses incorporating kinematic and 
inertial interaction may be considered for the estimation of seismic earth pressures on critical walls. 
Computer programs that may be utilized for such analyses include FLUSH (Lysmer et. al, 1975) and 
SASSI (Lysmer et al., 1981). 
Ostadan and White (1998) have observed that for embedded structures subjected to ground shaking, the 
characteristics of the wall pressure amplitudes vs. frequency of the ground motion were those of a 
single-degree-of-freedom (SDOF) system and proposed a simplified method to estimate the magnitude 
2003 Commentary, Chapter 7 
158 
and distribution of dynamic thrust. Results provided by Ostadan and White (1998) utilizing this 
simplified method, which were also confirmed by dynamic finite element analyses, indicate that, 
depending on the dynamic properties of the backfill as well as the frequency characteristics of the input 
ground motion, a range of dynamic earth pressure solutions would be obtained for which the M-O 
solution and the Wood (1973) solution represent a lower and an upper bound, respectively. 
Chang and others (1990) have found that dynamic earth pressures recorded on the wall of a model 
nuclear reactor containment building were consistent with dynamic pressures predicted by the M-O 
solution. Analysis by Chang and others indicated that the dynamic wall pressures were strongly 
correlated with the rocking response of the structure. Whitman (1991) has suggested that SSI effects on 
basement walls of buildings reduce dynamic earth pressures and that M-O pressures may be used in 
design except where structures are founded on rock or hard soil (that is, no significant rocking). In the 
latter case, the pressures given by the Wood (1973) formulation would appear to be more applicable. 
The effect of rocking in reducing the dynamic earth pressures on basement walls also has been 
suggested by Ostadan and White (1998). This condition may be explained if it is demonstrated that the 
dynamic displacements induced by kinematic and inertial components are out of phase. 
Effect of saturated backfill on wall pressures. The previous discussions are limited to backfills that are 
not water-saturated. In current (1999) practice, drains typically are incorporated in the design to prevent 
groundwater from building up within the backfill. This is not practical or feasible, however, for 
waterfront structures (such as quay walls) where most of the earthquake-induced failures have been 
reported (Seed and Whitman, 1970; Ebeling and Morrison, 1992; ASCE-TCLEE, 1998). 
During ground shaking, the presence of water in the pores of a backfill can influence the seismic loads 
that act on the wall in three ways (Ebeling and Morrison, 1992; Kramer, 1996): (1) by altering the 
inertial forces within the backfill, (2) by developing hydrodynamic pressures within the backfill and (3) 
by generating excess porewater pressure due to cyclic straining. Effects of the presence of water in 
cohesionless soil backfill on seismic wall pressures can be estimated using formulations presented by 
Ebeling and Morrison (1992). 
A soil liquefaction condition behind a wall may under the design earthquake have a pronounced effect 
on the wall pressures during and for some time after the earthquake. At present (1999), there is no 
general consensus established for estimating lateral earth pressures for liquefied backfill conditions. 
One simplified and probably somewhat conservative approach is to treat the liquefied backfill as a 
heavy viscous fluid exerting a hydrostatic pressure on the wall. In this case, the viscous fluid has the 
total unit weight of the liquefied soil. If unsaturated soil is present above the liquefied soil, it is treated 
as a surcharge that increases the fluid pressure within the underlying liquid soil by an amount equal to 
the thickness times the total unit weight of the surcharge soil. In addition to these static fluid 
pressures exerted by a liquefied backfill, hydrodynamic pressures can be exerted by the backfill. The 
magnitude of any such hydrodynamic pressures would depend on the level of shaking following 
liquefaction. Hydrodynamic effects may be estimated using formulations presented by Ebeling and 
Morrison (1992). 
7.5.3 Foundation ties. The additional requirement is made that spread footings on soft soil profiles 
should be interconnected by ties. The reasoning explained above under Sec. 7.4.3 also applies here. 
7.5.4 Special pile and grade beam requirements. For Seismic Design Categories D, E, and F, 
additional pile reinforcement over that specified for Seismic Design Category C buildings is required. 
Adequate pile ductility is required and provision must be made for additional reinforcement to ensure, 
as a minimum, full ductility in the upper portion of the pile. 
Special consideration is required in the design of concrete piles subject to significant bending during 
earthquake shaking. Bending can become crucial to pile design where portions of the foundation piles 
are supported in soils such as loose granular materials and/or soft soils that are susceptible to large 
deformations and/or strength degradation. Severe pile bending problems may result from various 
combinations of soil conditions during strong ground shaking, for example: 
Foundation Design Requirements 
159 
1. Soil settlement at the pile-cap interface either from consolidation of soft soil prior to the earthquake 
or from soil compaction during the earthquake can create a free-standing short column adjacent to 
the pile cap. 
2. Large deformations and/or reduction in strength resulting from liquefaction of loose granular 
materials can cause bending and/or conditions of free-standing columns. 
3. Large deformations in soft soils can cause varying degrees of pile bending. The degree of pile 
bending will depend upon thickness and strength of the soft soil layer(s) and/or the properties of the 
soft/stiff soil interface(s). 
Such conditions can produce shears and/or curvatures in piles that may exceed the bending capacity of 
conventionally designed piles and result in severe damage. Analysis techniques to evaluate pile bending 
are discussed by Margason and Holloway (1977) and Mylonakis (2001) and these effects on concrete 
piles are further discussed by Shepard (1983). For homogeneous, elastic media and assuming the pile 
follows the soil, the free-field curvature (soil strains without a pile present) can be estimated by dividing 
the peak ground acceleration by the square of the shear wave velocity of the soil, although considerable 
judgment is necessary in utilizing this simple relationship in a layered, inelastic profile with pile-soil 
interaction effects. Norris (1994) discusses methods to assess pile-soil interaction with regard to pile 
foundation behavior. 
The designer needs to consider the variation in soil conditions and driven pile lengths in providing for 
pile ductility at potential high curvature interfaces. Interaction between the geotechnical and structural 
engineers is essential. 
It is prudent to design piles to remain functional during and following earthquakes in view of the fact 
that it is difficult to repair foundation damage. The desired foundation performance can be 
accomplished by proper selection and detailing of the pile foundation system. Such design should 
accommodate bending from both reaction to the buildings inertial loads and those induced by the 
motions of the soils themselves. Examples of designs of concrete piles include: 
1. Use of a heavy spiral reinforcement and 
2. Use of exterior steel liners to confine the concrete in the zones with large curvatures or shear 
stresses. 
These provide proper confinement to ensure adequate ductility and maintenance of functionality of the 
confined core of the pile during and after the earthquake. 
Design of piles incorporates the same R force reduction factor as the superstructure and therefore 
implies inelasticity in the foundations and piles. Therefore, piles should be designed with similar 
ductility requirements as the superstructure. Foundations in SDC D, E, and F are expected to 
experience strong ground motions and large pile curvatures. Inertial pile-soil-structure interaction may 
produce plastic hinging in the piles near the bottom of the pile cap. Kinematic soil-pile-structure 
interaction will result in some bending moments and shearing forces throughout the length of the pile 
and will be higher at interfaces between stiff and soft soil strata. Inertial pile-soil-structure interaction 
will be particularly severe in soft soils and liquefiable soils located near the pile cap. This could result in 
plastic hinging of the pile in reverse curvature near the pile cap and for this reason the potential plastic 
hinge region is extended to seven pile diameters from the pile cap in the Provisions. 
Precast prestressed concrete piles are exempted from the concrete special moment frame detailing 
requirements adapted for concrete piles since these provisions were never intended for slender precast 
prestressed concrete elements and will result in unbuildable piles. Piles with substantially less 
confinement reinforcement than required by ACI 318 equation 10-6 have been proven through cyclic 
testing to have adequate performance (Park and Hoat Joen, 1990). Transverse steel requirements for the 
precast prestressed concrete piles are given in Section 7.5.4.4. 
Where grade beams have the strength to resist the load combination with overstrength, which simulates 
2003 Commentary, Chapter 7 
160 
expected foundation demands under a yielding structure, detailing similar to the beams of the Special 
Moment Frame is not required. This strong grade beam provision could apply to both cantilever 
column systems and frame systems with the objective of avoiding the inelastic response or plastic 
hinging of the grade beam where it would be difficult to detect and repair after being subjected to strong 
ground motions. 
Anchorage of the pile to the pile cap should be designed to permit energy dissipating mechanisms, such 
as pile slip at the pile-soil-interface, while maintaining a competent connection to the pile cap. A least 
capacity design approach is used for this purpose based on the pile section strength, not to exceed the 
load combination with overstrength, which simulates expected foundation demands under a yielding 
structure. A similar approach is also used for pile splice design. 
Provisions are given to establish requirements as to when different pile analysis methods should be 
used. Short piles and long slender piles embedded in the earth behave differently when subject to lateral 
forces and displacements. Long slender pile response depends on its interaction with the soil 
considering the non-linear response of the soil. Long slender piles should be analyzed for lateral loads 
considering the nonlinear interaction of the shaft and soil. The nonlinearity is typically considered in 
the soil and not the pile. Numerous design aid curves and computer programs are available for this type 
of analysis, and such an analysis is not uncommon in practice (e.g. Ensoft, 2000a). This type of analysis 
is necessary to obtain realistic pile moments, forces and deflections. More sophisticated analyses which 
also consider nonlinear behavior or plastic hinging in the pile itself as well as nonlinearity in the soil for 
actual earthquake ground motions is still in the research realm. For pile length-to-diameter ratios less 
than or equal to 6, the pile can be treated as a rigid body simplifying the analysis. A method assuming a 
rigid body and linear soil response for lateral bearing is in current building codes. A more accurate and 
comprehensive approach using this method is given in a study by Czerniak (1957). 
The effects of groups of piles, where closely spaced, should be taken into account for the soil vertical 
and horizontal response. As groups of closely spaced piles move laterally, failure zones for individual 
piles overlap and horizontal strength and stiffness response of the pile-soil system is reduced. 
Reduction factors or p-multipliers are needed to account for these groups of closely spaced piles. For 
a pile center-to-center spacing of three pile diameters, reduction factors of 0.6 for the leading pile row 
and 0.4 for the trailing pile rows are recommended by Rollins et al. (1999). Computer programs are 
available to analyze group effects, (e.g. Ensoft, 2000b). 
Batter pile systems that are partially embedded have historically performed poorly under strong ground 
motions (Gerwick and Fotinos, 1992). Failure of battered piles has been attributed to neglecting the 
potential loading on the piles from ground deformation and also to an erroneous assumption that the 
lateral loads will be resisted by the axial response of piles leading to neglect of the induced moments in 
the pile at the pile cap (Lam and Bertero, 1990). Difficulties in examining fully embedded batter piles 
have led to uncertainties of the extent of damage for this type of foundation. Batter piles are considered 
as limited ductile systems by their nature and should be designed using the load combination with 
overstrength. Due to eccentricities inherent in batter pile configurations, moment resisting connections 
to the pile cap are required to resolve the statics. Otherwise the superstructure will have to resolve the 
eccentricities by resisting moments induced by the foundation under lateral forces. This concept is 
clearly illustrated in EQE Engineering (1991). 
7.5.4.1 Uncased concrete piles. The uncased concrete piles category has been expanded to include 
auger-cast piles which are now subject to the same reinforcement requirements as other cast-in-place 
concrete piles. The longitudinal reinforcement within a pile has prescriptive termination point 
minimums intended for firm soils and is required to extend at least the full flexural length. The 
longitudinal reinforcement should extend past the flexural length by its development length 
requirement. The flexural length has been defined as that length of a pile from bottom of pile cap to a 
point where 0.4 times the concrete section cracking moment exceeds the calculated flexural demand. 
The 0.4 factor is analogous to a material resistance factor in strength design. This definition implies the 
plain concrete section will be resisting some minimal amount of moment demand beyond the flexural 
length. Where the pile is subject to significant uplift forces, it is recommended that the longitudinal 
Foundation Design Requirements 
161 
reinforcement extend the full length of the pile. 
Increased transverse reinforcement requirements are given for the potential plastic hinge zone 
immediately below the pile cap and for regions beyond that zone. The potential plastic hinge zone was 
taken to be three pile diameters below the pile cap to allow for the varied soil site classes from A 
through D. For soil site classes E and F, the potential plastic hinge zone is taken to be seven diameters in 
length as given in Section 7.5.4. The transverse reinforcement detailing for these zones is similar to that 
required for concrete Special Moment Frames at hinge regions and is expected to provide a 
displacement ductility of approximately 5 to 6 depending upon the axial load. However, recent studies 
and testing has substantiated that the soil will contribute substantially to the confinement of the concrete 
pile section in firm soils. Chai and Hutchison (1998) found that in-situ lateral load testing of 16 inch 
diameter conventionally reinforced circular piles performed to displacement ductilities from 
approximately 3 to 4 using a spiral steel reinforcement ratio of 0.0057. Further testing of the same piles 
but with a spiral steel reinforcement ratio of 0.0106 produced improved displacement ductilities and no 
spiral fracture failure which occurred in the prior piles tested with the lower spiral ratio. These circular 
spiral ratios are considerably less than those required by ACI 318 for columns in Seismic Design 
Category D. ACI 318 equation 10-6 would require a spiral reinforcement ratio of at least 0.0175 
depending on the concrete core diameter. Budek, Benzoni and Priestley (1997) found that testing of 24 
inch diameter circular conventionally reinforced concrete piles with a transverse (circular) 
reinforcement ratio of 0.006 offered adequate performance up to a displacement ductility of 4. Their 
conclusions indicate that the soil confinement can play a significant role in the pile shaft response As a 
result of these studies, full confinement reinforcement intended for superstructure columns is not 
necessary for in-ground pile foundations, except for soil site classes E and F and liquefiable soils. 
Beyond the potential plastic hinge region, the curvature ductility demand is not considered to exceed 
that provided by the nominal moment capacity of the section for nominal earthquake loads. 
7.5.4.4 Precast-prestressed piles. The transverse reinforcement requirements are primarily taken from 
the PCI Committee Report (1993) on precast prestressed concrete piling for geographic regions subject 
to strong ground motions. The circular spiral transverse reinforcement equations recommended for 
precast prestressed concrete piles given in Park and Hoat Joen (1990) are the basis for the provisions. 
These equations make the curvature ductility capacity dependent on the pile axial load. A reduction of 
50% in the normally required circular spiral reinforcement from those equations (similar to ACI 318 
equation 10-6) was sufficient to achieve a displacement ductility of 4. The reduced circular spiral 
transverse reinforcement requirement is the basis for the PCI Piling Committees final provisions. 
High strength hard drawn wire with higher yield strengths is permitted to be used for transverse circular 
spiral reinforcement of precast prestressed concrete piles. Pile test specimens using this type of 
transverse reinforcement include the research done by Park and Hoat Joen (1990). High strength hard 
drawn wire has yield strengths between 150 and 200 ksi. fyh is conservatively limited to 85 ksi for this 
steel because hard drawn wire has limited ductililty. 
7.5.4.5 Steel Piles. AISC Seismic (2002), Part I, Section 8.6 provides seismic design and detailing 
provisions for steel H-piles which should be used in conjunction with these provisions. 
REFERENCES 
AISC, 2002, Seismic Provisions for Structural Steel Buildings, American Institute of Steel 
Construction, Chicago, Illinois. 
American Society of Civil Engineers, Technical Committee on Lifeline Earthquake Engineering. 1998. 
Seismic Guidelines for Ports, edited by S. D. Werner, Monograph 12. 
Bardet, J.P., Tobita, T., Mace, N., and Hu, J., 2002, Regional modeling of liquefaction-induced ground 
deformation, Earthquake Spectra, 18(1), 19-46. 
Blake, T.F., Hollingsworth, R.A., and Stewart, J.P., editors 2002, Recommended procedures for 
implementation of DMG Special Publication 117 Guidelines for analyzing and mitigating landslide 
2003 Commentary, Chapter 7 
162 
hazards in California, Southern California Earthquake Center, University of Southern California, Los 
Angeles, California, 130 p. 
Bray, J.D., and Rathje, E.M. 1998, Earthquake-induced displacements of solid-waste landfills, Journal 
of Geotechnical and Geoenvironmental Engrg., ASCE, 124, 242-253. 
Budek, A., Benzoni, G. and Priestley, M.J.N., 1997, The Effect of External Confinement and 
Transverse Reinforcement on Plastic Hinges in Cast-in-place Pile Shafts - Experimental Results, 
Report No. SSRP - 97/04, University of California, San Diego. 
California Department of Conservation, Division of Mines and Geology, 1997, Guidelines for 
Evaluating and Mitigating Seismic Hazards in California. 
California Geological Survey, 2002, Guidelines for evaluating the hazard of surface fault rupture: 
California Geological Survey Note 49, 4 p. 
Chai, Y. H. and Hutchison, T. C., Large Scale Lateral Load Tests of Reinforced Concrete Bridge Piles 
in Cohesionless Soil, SEAOC 1998 Convention Proceedings. 
Chang, C. Y., M. S. Power, C. M. Mok, Y. K. Tang, and H. T. Tang. 1990. Analysis of Dynamic 
Lateral Earth Pressures Recorded on Lotung Reactor Containment Model Structure, in Proceedings, 
Fourth U.S. National Conference on Earthquake Engineering, pp. 643-652. EERI. 
Czerniak, E., "Resistance to Overturning of Single Short Piles", 1957, ASCE Journal of the Structural 
Division, Vol 83, ST-2. 
Das, B. M. 1983. Fundamentals of Soil Dynamics. New York: Elsvier. 
Ebeling, R. M., and E. E. Morrison. 1992. The Seismic Design of Waterfront Retaining Structures, 
Technical Report ITL-92-11. Vicksburg, Mississippi: Corps of Engineers Waterways Experiment 
Station. 
Ensoft, Inc., 2000b, Computer Program GROUP, Version 5.0, A Program for the Analysis of a Group of 
Piles Subjected to Axial and Lateral Loading, Users Manual and Technical Manual, Austin, Texas. 
Ensoft, Inc., 2000a, Computer Program LPILE Plus Version 4.0, A Program for the Analysis of Piles 
and Drilled Shafts under Lateral Loads, Users Manual and Technical Manual, Austin, Texas. 
EQE Engineering ,1991 Structural Concepts and Details for Seismic Design, UCRL-CR-106554, 
Dept. of Energy. 
Franklin, A.G. and Chang, F.K., 1977, "Earthquake Resistance of Earth and Rock-Fill Dams; Permanent 
Displacements of Earth Embankments by Newmark Sliding Block Analysis," Miscellaneous Paper S- 
71-17, Report 5, U.S. Army Waterways Experiment Station, CE, Vicksburg, MS. 
Gerwick, Jr., B., Fotinos, G., 1992, "Drilled Piers and Driven Piles for Foundations in Areas of High 
Seismicity", SEAONC Fall Seminar, October 29. 
Goodman, R.E. and Seed, H.B., 1966, Earthquake-induced displacements in Sand embankments, ASCE 
Journal of Soil Mechanics and Foundation, v. 92, no. SM2, p. 125-146.Hart, E.W., and Bryant, W.A., 
1997 (revised), Fault-rupture hazard zones in California: California Department of Conservation, 
Division of Mines and Geology (now California geological Survey) Special Publication 42, 38 p. 
Hart, E.W., and Bryant, W.A., 1997 (revised), Fault-rupture hazard zones in California: California 
Department of Conservation, Division of Mines and Geology (now California Geological Survey) 
Special Publication 42, 38p. 
Hynes, M.E. and Franklin, A.G., 1984, "Rationalizing the Seismic Coefficient Method," Miscellaneous 
Paper GL-84-13, U.S. Army Waterways Experiment Stations, Vicksburg, MS, July, 21 p. 
Idriss, I. M. 1985 Evaluating Seismic Risk in Engineering Practice. In Proceedings of the 11th 
International Conference on Soil Mechanics and Foundation Engineering, Vol. 1, p. 155. 
Foundation Design Requirements 
163 
Ishihara, Kenji. 1985. Stability of Natural Deposits During Earthquakes. In Proceedings of the 11th 
International Conference on Soil Mechanics and Foundation Engineering, San Francisco, Vol. 1, pp. 
321-376. 
Kelson, K.I., Kang, K.-H., Page, W.D., Lee, C.-T., and Cluff, L.S., 2001, Representative styles of 
deformation along the Chelungpu fault from the 1999 Chi-Chi (Taiwan) Earthquake: Geomorphic 
characteristics and responses of man-made structures: Bulletin of the Seismological Society of 
America, v. 91, no. 5, p. 930-952. 
Kramer, S. L. 1996. Geotechnical Earthquake Engineering. New Jersey: Prentice Hall. 
Lam, I., and G. R. Martin. 1986. Seismic Design of Highway Bridge FoundationsDesign Procedures 
and Guidelines, Report FHWA/RD-86/102. Washington, D.C.: FHWA. 
Lam, I., Bertero, V., 1990 Aseismic Design of Pile Foundations for Port Facilities, Proceedings of the 
POLA Seismic Workshop on Seismic Engineering, March 21-23, San Pedro, California, Port of Los 
Angeles. 
Lysmer, J., M. Tabatabai-Rassi, F. Jajirian, S. Vahdani, and F. Ostadan. 1981. SASSI C A System for 
Analysis of Soil-Structure Interaction, Report UCB/GT/81-02. Berkeley: University of California 
Department of Engineering. 
Lysmer, J., T. Udka, C. P. Tsai, and H. B. Seed. 1975. FLUSH: A Computer Program for 
Approximate 2-D Analysis of Soil-Structure Interaction Problems, Report EERC 75-30. Berkeley: 
University of California, Earthquake Engineering Research Center. 
Makdisi, F. T., and H. B. Seed. 1978. Simplified Procedure for Estimating Dam and Embankment 
Earthquake-Induced Deformations. Journal of Geotechnical Engineering ASCE 104(GT7):849-867. 
Margason, E., and M. Holloway. 1977. Pile Bending During Earthquakes. In Proceedings of the 6th 
World Conference on Earthquake Engineering, New Delhi. 
Martin, G.R., and Qiu, P., 1994, "Effects of Liquefaction on Vulnerability Assessment," NCEER 
Highway Project on Seismic Vulnerability of New and Existing Highway Construction, One Year 
Research Tasks  Technical Research Papers, 1994. 
Martin, G.R., and Lew, M. 1999, "Recommended Procedures for Implementation of DMG Special 
Technical Publication 117, Guidelines for Analyzing and Mitigating Liquefaction in California," 
Southern California Earthquake Center, University of Southern California, March, 63 p. 
Martin, G.R., and Lam, I.P., 2000, Earthquake Resistant Design of Foundations: Retrofit of Existing 
Foundations, Proceedings, GeoEng 2000, International Conference on Geological and Geotechnical 
Engineering, Melbourne, Australia, November. 
Monobe, N., and H. Matsuo. 1929. On the Determination of Earth Pressures During Earthquakes. 
In Proceedings, World Engineering Congress. 
Mylonakis, G., 2001, Seismic pile bending at soil-layer interfaces, Soils and Foundations, 41 (4), 47-58. 
Nadim, F. 1980. Tilting and Sliding of Gravity Retaining Walls, MS thesis, Department of Civil 
Engineering, Massachusetts Institute of Technology, Cambridge. 
Nadim, F., and R. V. Whitman. 1984. Coupled Sliding and Tilting of Gravity Retaining Walls During 
Earthquakes. In Proceedings of the Eighth World Conference on Earthquake Engineering, pp. 477- 
484. National Research Council, Committee on Earthquake Engineering. 1985. Liquefaction of Soils 
During Earthquakes. Washington, D.C.: National Academy Press. 
National Center for Earthquake Engineering Research (NCEER). 1997. Proceedings of the NCEER 
Workshop on Evaluation of Liquefaction Resistance of Soils, Technical Report NCEER-97-0022, 276 
p. 
2003 Commentary, Chapter 7 
164 
National Research Council. 1985. Liquefaction of Soils During Earthquakes, National Academy Press, 
240p. 
Newmark, N.M. 1965. Effects of earthquakes on dams and embankments, Geotechnique, London, 
England v.5, no.2. 
Norris, G. M. 1994. Seismic Bridge Pile Foundation Behavior, in Proceedings, International 
Conference on Design and Construction of Deep Foundations, FHWA, Vol. 1. 
Okabe, S. 1926. General Theory of Earth Pressure, Journal of the Japan Society of Civil Engineers 
12,1. 
Ostadan, F., and W. H. White. 1998. Lateral Seismic Soil PressureAn Updated Approach. In 
Preproceedings of UJNR Workshop on Soil-Structures Interaction, U.S. Geological Survey, Menlo 
Park, California. 
Park, R. and Hoat Joen, P., 1990. Simulated Seismic Load Tests on Prestressed Concrete Piles and 
Pile-Pile Cap Connections, PCI Journal, November-December. 
Prakash, S. 1981. Soil Dynamics. New York: McGraw-Hill Book Co. 
Prakash, S., Y. Wu, and E. A. Rafnsson. 1995. Displacement Based Aseismic Design Charts for Rigid 
Walls. Rolla, Missouri: Shamsher Prakash Foundation. 
Precast Concrete Institute, 1993. Recommended Practice for Design, Manufacture and Installation of 
Prestressed Concrete Piling, PCI Committee on Pre-stressed Concrete Piling, PCI Journal, March- 
April. 
Richards, R. J., and D. Elms. 1979. Seismic Behavior of Gravity Retaining Walls. Journal of the 
Geotechnical Engineering Divisions, ASCE 105(GT4): 449-464. 
Richards, R., K. L. Fishman, and R. Devito. 1996. Threshold Accelerations for Rotation or Sliding of 
Bridge Abutments, Journal of Geotechnical Engineering, ASCE 105, GT4, 449-464. 
Robertson, P.K., and Wride, C.E. 1998. Evaluation of cyclic liquefaction potential using the cone 
penetration test, Canadian Geotechnical Journal, Ottawa, 35(3), 442-459. 
Rollins, Kyle M., Peterson, KT., Weaver, T.J., Sparks, Andrew E., 1999. Static and dynamic lateral load 
behavior on a full-scale pile group in clay, Brigham Young University, Provo, Utah and the Utah 
Department of Transportation, Salt Lake City, Utah, 23 June. 
Seed, H. B., and I. M. Idriss. 1971. Simplified Procedure for Evaluating Soil Liquefaction Potential. 
Journal of the ASCE Soil Mechanics and Foundations Division 97(SM9):1249-1273. 
Seed, H. B., K. Tokimatsu, L. F. Harder, and R. M. Chung. 1985. The Influence of SPT Procedures in 
Soil Liquefaction Resistance Evaluations, Report UBC/EERC-84/15. Berkeley, California: Earthquake 
Engineering Research Center. 
Seed, H. B., I. M. Idriss, and I. Arango. 1983. Evaluation of Liquefaction Potential Using Field 
Performance Data. ASCE Journal of Geotechnical Engineering 109(3):458-482. 
Seed, H. B., and I. M. Idriss. 1982. Ground Motions and Soil Liquefaction During Earthquakes, Vol. 5 
of Engineering Monographs on Earthquake Criteria, Structural Design, and Strong Motion Records. 
El Cerrito, California: Earthquake Engineering Research Institute. 
Seed, R. B., and L. F. Harder. 1990. SPT-Based Analysis of Cyclic Pore Pressure Generation and 
Undrained Residual Strength. In Proceedings of the H. Bolton Seed Memorial Symposium, Vol. 2, 
pp. 351-376. 
Seed, H. B., and R. V. Whitman. 1970. Design of Earth Retaining Structures for Dynamic Loads, in 
Proceedings, ASCE Specialty Conference on Lateral Stresses in the Ground and Design of Earth- 
Retaining Structures, pp. 103-147. Ithaca, New York: Cornell University. 
Shepard, D. A. 1983. Seismic Design of Concrete Piling. PCI Journal (March/April). 
Foundation Design Requirements 
165 
Siddarthan, R., S. Ara, and G. Norris. 1992. Simple Rigid Plastic Model for Seismic Tilting of Rigid 
Walls, Journal of Structural Engineering, ASCE, 469-487. 
Slemmons, D.B., and dePolo, C.M., 1986, Evaluation of active faulting and associated hazards, in 
Studies in GeophysicsActive Tectonics: National Academy Press, Washington, D.C., p. 45-62. 
Soydemir, C., and M. Celebi. 1992. Seismic Design of Buildings with Multi-Level Basements, in 
Proceedings, Tenth World Conference on Earthquake Engineering, Madrid Spain, Vol. 6, pp. 1731- 
1734. 
Stark, T.D., and Mesri, G., 1992. Undrained Shear Strength of Liquefied Sands for Stability Analysis, 
Journal of Geotechnical Engineering, American Society of Civil Engineers, v. 118, no. 11, p. 1727- 
1747. 
Steedman, R. S., and X. Zeng. 1996. Rotation of Large Gravity Retaining Walls on Rigid 
Foundations Under Seismic Loading. In Analysis and Design of Retaining Walls Against Earthquakes, 
ASCE Geotechnical Special Publication 60, edited by S. Prakash, pp. 38-56. New York: ASCE. 
Stewart, J.P., Blake, T.M., and Hollingsworth, R.A., 2003, A screen analysis procedure for seismic 
slope stability: Earthquake Spectra, 19 (3). 
Swan, F.H., Stepp, J.C., McGuire, R.K., 1991, Assessment of the potential for tectonic fault rupture for 
high-level nuclear waste repositories: Quarterly Journal of Engineering Geology, v. 24, p. 337-346. 
Tokimatsu, K., and H. B. Seed. 1987. Evaluation of Settlements in Sands Due to Earthquake 
Shaking. Journal of Geotechnical Engineering ASCE 113(8):861-878. 
U.S. Army Corps of Engineers, 1988. Seismic Design for Buildings, TI-809-04, Appendix F, Geologic 
Hazards Evaluations. 
Utah Section of the Association of Engineering Geologists, 1987, Guidelines for evaluating surface fault 
rupture hazard in Utah: Utah Geological and Mineral Survey Miscellaneous Publication N, 2 p. 
Veletsos, A., and A. H. Younan. 1994a. Dynamic Soil Pressure on Rigid Vertical Walls, Earthquake 
Engineering and Soil Dynamics, 23:275-301. 
Veletsos, A. and Younan, A. H. 1994b. Dynamic Modeling and Response of Soil-Wall Systems, 
Journal of Geotechnical Engineering, ASCE, 120:12. 
Wells, D.L., and Coppersmith, K.J., 1994, "New Empirical Relationships Among Magnitude, Rupture 
Area, and Surface Displacement," Bulletin of the Seismological Society of America, Vol, 84, p. 974- 
1002. 
Whitman, R. V. 1991. Seismic Design of Earth Retaining Structures, in Proceedings, Second 
International Conference on Recent Advances in Geotechnical Earthquake Engineering and Soil 
Dynamics, St. Louis, Missouri. 
Whitman, R. V., and S. Liao. 1985. Seismic Design of Retaining Walls, Miscellaneous Paper GL-85-1. 
Vicksburg, Mississippi: Corps of Engineers Waterways Experiment Station. 
Wong, C.P. and Whitman, R.V., 1982. "Seismic Analysis and Improved Seismic Design Procedure for 
Gravity Retaining Walls," Research Report 82-83, Department of Civil Engineering, Massachusetts 
Institute of Technology, Cambridge, MA. 
Wood, J. H. 1973. Earthquake-Induced Soil Pressures on Structures, Report EERL 73-05. Pasadena: 
California Institute of Technology. 
Youd, T. L. 1989. Ground Failure Damage to Buildings During Earthquakes. In Foundation 
EngineeringCurrent Principles and Practices, Vol. 1, pp. 758-770. New York: American Society of 
Civil Engineers. 
2003 Commentary, Chapter 7 
166 
Youd, T.L., and Noble, S.K., 1997, Magnitude scaling factors, Proc. NCEER Workshop on Evaluation 
of Liquefaction Resistance of Soils, National Center for Earthquake Engineering Research, State 
University of New York at Buffalo, 149-165. 
Youd, T.L., Hansen, C.M., and Bartlett, S.F., 2002, Revised multilinear repression equations for 
prediction of lateral spread displacement, ASCE Journal of Geotechnical and Geoenvironmental 
Engineering, 128:12, p. 1007-1017. 
Youd, T.L., Idriss, I.M., Andrus, R.D., Arango, I., Castro, G., Christian, J.T., Dobry, R., Finn, W., D.L. 
Harder, L.F., Jr., Hynes, M.E., Ishihara, K., Koester, J.P., Liao, S.S.C., Marcuson, W. F., III, Martin, 
G.R. Mitchell, J.K., Moriwaki, Y., Power, M.S., Robertson, P.K., Seed, R.B., and Stokoe, K.H., II, 
2001. Liquefaction resistance of soils: Summary Report from the 1996 NCEER and 1998 NCEER/NSF 
Workshops on Evaluation of Liquefaction Resistance of Soils, Journal of Geotechnical Geoenvironmental 
Engineering, ASCE, 127 (10), p. 817-833. 
Youngs, R.R., Arabasz, W.J., Anderson, R.E., Ramelli, A.R., Ake, J.P., Slemmons, D.B., McCalpin, 
J.P., Doser, D.I., Fridrich, C.J., Swan III, F.H., Rogers, A.M., Yount, J.C., Anderson, L.W., Smith, 
K.D., Bruhn, R.L., Kneupfer, P.L.K., Smith, R.B., dePolo, C.M., OLeary, D.W., Coppersmith, K.J., 
Pezzopane, S.K., Schwartz, D.P., Whitney, J.W., Olig, S.S., and Toro, G.R., 2003, A methology for 
probabilistic fault displacement hazard analysis (PFDHA): Earthquake Spectra, v. 19, no. 1, p. 191-219. 

 167 
Appendix to Chapter 7 
GEOTECHNICAL ULTIMATE STRENGTH DESIGN OF FOUNDATIONS AND 
FOUNDATION LOAD-DEFORMATION MODELING 
A7.2 GENERAL DESIGN REQUIREMENTS 
A7.2.2. Foundation load capacities. In current geotechnical engineering practice, foundation design is 
based on allowable stresses, with allowable foundation load capacities, Qas, for dead plus live loads 
based on limiting static settlements and providing a large factor of safety against exceeding ultimate 
capacities. In current practice, allowable soil stresses for dead plus live loads are increased by one-third 
for load combinations that include wind or seismic forces. The one-third increase is overly conservative 
if the allowable stresses for dead plus live loads are far below ultimate soil capacity. 
This appendix provides guidance for the direct use of ultimate foundation load capacity, Qus, for load 
combinations including seismic effects. It is required that foundations be capable of resisting loads with 
acceptable deformations considering the short duration of seismic loading, the dynamic properties of the 
soil, and the ultimate load capacities, Qus, of the foundations under vertical, lateral, and rocking loading. 
A7.2.2.1. Determination of ultimate foundation load capacities. For competent soils that are not 
expected to degrade in strength during seismic loading (e.g., due to partial or total liquefaction of 
cohesionless soils or strength reduction of sensitive clays), use of static soil strengths is recommended 
for determining ultimate foundation load capacities, Qus. Use of static strengths is somewhat 
conservative for such soils because rate-of-loading effects tend to increase soil strengths for transient 
loading. Such rate effects are neglected because they may not result in significant strength increase for 
some soil types and are difficult to confidently estimate without special dynamic testing programs. The 
assessment of the potential for soil liquefaction or other mechanisms for reducing soil strengths is 
critical, because these effects may reduce soil strengths greatly below static strengths in susceptible 
soils. 
The best-estimated ultimate vertical load capacity of footings, Qus, should be determined using accepted 
foundation engineering practice. In the absence of moment loading, the ultimate vertical load capacity 
of a rectangular footing of width B and length L may be written as 
Qus = qcBL 
where qc = ultimate soil bearing pressure. 
For rigid footings subject to moment and vertical load, contact stresses become concentrated at footing 
edges, particularly as footing uplift occurs. The ultimate moment capacity, Mus, of the footing as limited 
by the soil is dependent upon the ratio of the vertical load stress, q, to the ultimate soil bearing pressure 
qc. Assuming that contact stresses are proportional to vertical displacements and remain elastic up to qc, 
it can be shown that uplift will occur prior to plastic yielding of the soils when q/qc is less than 0.5. If q/ 
qc is greater than 0.5, then the soil at the toe will yield prior to uplift. This is illustrated in Figure 
CA7.2.2-1. In general the ultimate moment capacity of a rectangular footing may be expressed as: 
. .. 
. 
. .. 
. 
= - 
qc 
LP q 
M us 1 
2 
where 
2003 Commentary, Appendix to Chapter 7 
168 
P = Vertical Load 
q = 
BL 
P 
B = Footing width, and 
L = Footing length in direction of rotation 
The ultimate lateral load capacity of a footing may be assumed equal to the sum of the best-estimated 
ultimate soil passive resistance against the vertical face of the footing plus the best-estimated ultimate 
soil friction force on the footing base. The determination of ultimate passive resistance should consider 
the potential contribution of friction on the face of the footing on the passive resistance. 
For piles, the best-estimated ultimate vertical load capacity (for both axial compression and axial tensile 
loading) should be determined using accepted foundation engineering practice. When evaluating axial 
tensile load capacity, consideration should be given to the capability of pile cap and splice connections 
to take tensile loads. 
The ultimate moment capacity of a pile group should be determined assuming a rigid pile cap, leading to 
an initial triangular distribution of axial pile loading from applied overturning moments. However, full 
axial capacity of piles may be mobilized when computing ultimate moment capacity, in a manner 
analogous to that described for a footing in Figure CA7.2.2-1. The ultimate lateral capacity of a pile 
group may be assumed equal to the best-estimated ultimate passive resistance acting against the edge of 
the pile cap and the additional passive resistance provided by piles. 
Resistance factors, f, are provided to factor ultimate foundation load capacities, Qus, to reduced 
capacities, fQus, used to check foundation acceptance criteria. The values of f recommended in the 
Provisions are higher than those recommended in some codes and specifications for long-term static 
loading. The development of resistance factors for static loading has been based on detailed reliability 
studies and on calibrations to give designs and factors of safety comparable to those given by allowable 
stress design. As indicated in the first paragraph of this section, mobilized strengths for seismic loading 
conditions are expected to be somewhat higher than the static strengths specified for use in obtaining 
values of Qus, especially for cohesive soils. In the absence of any detailed reliability studies for seismic 
loading conditions, Values of f equal to 0.8 and 0.7 were selected for cohesive and cohesionless soils, 
respectively, where geotechnical site investigations, including laboratory or insitu tests, are conducted, 
Figure CA7.2.2-1. 
Geotechnical Ultimate Strength Design of Foundations and Foundation Load-deformation Modeling 
169 
and values of f equal to 1.0 and 0.9 were selected where full-scale field testing of prototype foundations 
are conducted. These values are comparable to the values of 0.8 (for soil strengths determined based on 
a comprehensive site soil investigation including soil sampling and testing) and 0.9 (for soil strengths 
determined by site loading testing using plate bearing or near full scale foundation element testing) 
recommended by the SEAOC Seismology Committee Ad Hoc Foundation Committee (2001). 
A7.2.2.2 Acceptance criteria. The factored load capacity, fQus, provides the basis for the acceptance 
criteria, particularly for the linear analysis procedures. The mobilization of ultimate capacity in the 
nonlinear analysis procedures does not necessarily mean unacceptable performance as structural 
deformations due to foundation displacements may be tolerable, as discussed by Martin and Lam 
(2000). For the nonlinear analysis procedures, it is also prudent to evaluate structural behavior utilizing 
parametric increases in foundation load capacities above Qus by a factor of 1/f, to check potential 
changes in structural ductility demands. 
A7.2.3 Foundation load-deformation modeling. Analysis methods described in Sec. 5.3 (response 
spectrum procedure) and Sec. 5.4 (linear response history procedure), permit the use of realistic 
assumptions for foundation stiffness, as opposed to the assumption of a fixed base. In addition, the 
nonlinear response history procedure (Sec. 5.5) and the nonlinear static procedure (Appendix to Chapter 
5) permit the use of realistic assumptions for the stiffness and load-carrying characteristics of the 
foundations. Guidance for flexible foundation (non-fixed base) modeling for the above analysis 
procedures are described herein. 
Foundation load-deformation behavior characterized by stiffness and load capacity may significantly 
influence the seismic performance of a structure, with respect to both load demands and distribution 
among structural elements (ATC 1996, NEHRP 1997a, 1997b). This is illustrated schematically in 
Figure CA 7.2.3-1. While it is recognized that the load-deformation behavior of foundations is 
nonlinear, an equivalent elasto-plastic representation of load-deformation behavior is often assumed, as 
illustrated in Figure CA 7.2.3-2. To allow for variability and uncertainty in the selection of soil 
parameters and analysis methods used to determine stiffness and capacity, a range of parameters for 
foundation modeling should be used to permit sensitivity evaluations. 

2003 Commentary, Appendix to Chapter 7 
170 
Figure CA7.2.3-1. 
Geotechnical Ultimate Strength Design of Foundations and Foundation Load-deformation Modeling 
171 
Figure CA7.2.3-2. 
Consider the spread footing shown in Figure CA 7.2.3-3 with an applied vertical load (P), lateral load 
(H), and moment (M). The soil characteristics might be modeled as two translational springs and a 
rotational spring, each characterized by a linear elastic-stiffness and a plastic capacity. The use of a 
Winkler spring model acting in conjunction with the foundation to eliminate the rotational spring may 
also be used, as shown in Figure CA7.2.3-4. The Winkler model can capture more accurately 
progressive mobilization of plastic capacity during rocking behavior. Note the lateral action is normally 
uncoupled from the vertical and rotational action. Many foundation systems are relatively stiff and 
strong in the horizontal direction, due to passive resistance against the face of footings or basement 
walls, and friction beneath footings and floor slabs. Comparisons of horizontal stiffness of the 
foundation and the structure can provide guidance on the need to include horizontal foundation stiffness 
in demand or capacity analyses. In general, foundation rocking has the most influence on structural 
response. Slender shear wall structures founded on strip footings, in particular, are most sensitive to the 
effects of foundation rocking. 
Assuming a shallow footing foundation may be represented by an embedded rigid plate in an elastic 
half-space, classical elastic solutions may be used to compute the uncoupled elastic stiffness parameters. 
Representative solutions are described in Commentary to Sec. 5.6. Solutions developed by Gazetas 
(1991) are also often used, as described in ATC (1996). Dynamic soil properties (i.e. properties 
consistent with seismic wave velocities and associated moduli of the soils as opposed to static soil 
moduli) should be used in dynamic soil solutions. The effects of nonlinearity on dynamic soil 
properties should be incorporated using the reduction factors in Sec. 5.6.2.1.1 or based on a site-specific 
study. 
Figure CA7.2.3-3. 
2003 Commentary, Appendix to Chapter 7 
172 
Figure CA 7.2.3-4. 
In the case of pile groups, the uncoupled spring model shown in Figure CA 7.2.3-3 may also be used, 
where the footing represents the pile cap. In the case of the vertical and rotational springs, it can be 
assumed that the contribution of the pile cap is relatively small compared to the contribution of the piles. 
In general, mobilization of passive pressures by either the pile caps or basement walls will control 
lateral spring stiffness. Hence, estimates of lateral spring stiffness can be computed using elastic 
solutions as for footings. In instances where piles may contribute significantly to lateral stiffness (i.e., 
very soft soils, battered piles), solutions using beam-column pile models are recommended. 
Axial pile group stiffness spring values, ksv, are generally in the range given by: 
ksv = 
L 
N AE 
n 
0.5 
1 
S
= 
to 
L
N AE 
n 
2 
1 
S
= 
where 
A = Cross-sectional area of a pile, 
E = Modulus of elasticity of piles, 
L = Length of piles , and 
N = Number of piles in group. 
Values of axial stiffness depend on complex nonlinear interaction of the pile and soil (NEHRP, 1997b). 
For simplicity, best estimate values of AE/L and 1.5 AE/L are recommended for piles where axial 
capacity is primarily controlled by end bearing and side friction, respectively. 
The rocking spring stiffness values, ksr, about each horizontal pile cap axis may be computed by 
assuming each axial pile spring acts as a discrete Winkler spring. The rotational spring constant 
(moment per unit rotation) is then given by: 
ksr = 
2 
1 
Sn 
N 
n 
S kvn 
= 

Geotechnical Ultimate Strength Design of Foundations and Foundation Load-deformation Modeling 
173 
where 
kvn = Axial stiffness of the nth pile, and 
Sn = Distance between nth pile and axis of rotation. 
The effects of group action and the influence of pile batter are not accounted for in the above equations. 
These effects should be evaluated if judged significant. 
Design Examples. In order to study and illustrate the effects of the change from allowable stress to ultimate 
strength design of foundations a series of examples were generated. The examples compared the size of 
foundations resulting from ultimate strength designs (USD) according the new procedures with those that would 
be obtained from conventional allowable stress designs (ASD). 
The examples were based upon a single six-story reinforced concrete building with shear walls and 
gravity frame (see Figure 7.2.3-5). One set of examples was for a shallow spread footing design 
beneath a shear wall. The other set applied to deep CIDH piers placed beneath the same wall. For each 
set of examples, individual designs reflected a range of soil strengths and ASD factors of safety. The 
vertical loads were not changed, but two levels of seismic overturning demand were imposed. 
Example 
CIDH piers 
Example spread footing 
Note: For a detailed description see ATC 40 vol.2 Example 
CIDH piers 
Example spread footing 
Example 
CIDH piers 
Example 
CIDH piers 
Example spread footing 
Note: For a detailed description see ATC 40 vol.2 
Figure 7.2.3-5 Example building. 
While is not possible to generalize the results of these examples to apply universally, they are 
representative of the effects of the change to USD for a realistic case study. For the spread footing 
foundation the area of the footing for USD compared to that for ASD is controlled by the factor of 
safety applied to the soil strength for vertical loads. This reduction ranged from 0% to 20 percent for a 
low FOS (2) up to 25 percent to 40 percent for a high FOS (4). This is not surprising; where ASD uses 
a high factor of safety and is thus most conservative, USD results in a smaller footing size. However, 
the footing size cannot be smaller than that required for allowable stresses for static design under 
vertical dead plus live loads. For the pier example, the required length for USD was actually about 50 
percent greater than for ASD for a low FOS (1.5) and up to 40 percent less for a high FOS (4). 
REFERENCES 
ATC, 1996, ATC 40, Seismic Evaluation and Retrofit of Concrete Buildings, 2 Volumes, The Applied 
Technology Council, Redwood City, CA. 
2003 Commentary, Appendix to Chapter 7 
174 
Gazetas, G. (1991), Foundation Vibrations, Foundation Engineering Handbook, 2nd Edition, Vas 
Nostrand Reinhold, Edited by Hsai-Yang Fang. 
Martin, G.R. and Lam, I.P., 2000, Earthquake Resistant Design of Foundations: Retrofit of Existing 
Foundations, Proceedings, GeoEng 2000, International Conference on Geological and Geotechnical 
Engineering, Melbourne, Australia, November. 
NEHRP, 1997a, and 1997b, National Earthquake Hazard Reduction Program Guidelines and 
Commentary for the Seismic Rehabilitation of Buildings, developed for the Building Seismic Safety 
Council for the Federal Emergency Management Agency, (Publications FEMA 273 and 274). 
SEAOC Seismology Committee, Ad Hoc Foundations Committee, 2001, USD/LRFD/Limit State 
Approach to Foundation Design, Proceedings of the 70th Annual SEAOC Convention, San Diego, 
California. 

 175 
Chapter 8 Commentary 
STEEL STRUCTURE DESIGN REQUIREMENTS 
8.1 GENERAL 
8.1.2 References. The reference documents presented in this section are the current specifications for 
the design of steel members, systems, and components in buildings as approved by the American 
Institute of Steel Construction (AISC), the American Iron and Steel Institute (AISI), the American 
Society of Civil Engineers (ASCE) and the Steel Joist Institute (SJI). 
Revise the AISC Seismic Commentary Sec. C9.3 as follows: At the end of the second paragraph add 
the following: This provision requires that the panel zone be proportioned using the method used to 
proportion the panel zone thickness of successfully tested connections. This should not be construed to 
mean that the thickness is required to be the same as the tested connection, only that the same method 
must be used to proportion it. For example, if the test were performed on a one-sided connection and 
the same beam and column sizes were used in a two-sided connection, the panel zone would be twice as 
thick as that of the tested connection. 
8.2 GENERAL DESIGN REQUIREMENTS 
8.2.1 Seismic Design Categories B and C. Structures assigned to Seismic Design Categories B and C 
do not require the same level of ductility capacity to provide the required performance as those assigned 
to the higher categories. For this reason, such structures are permitted to be designed using the 
requirements of any of the listed references, provided that the lower R value specified in Table 4.3-1 is 
used. Should the registered design professional choose to use the higher R values in the table, the 
detailing requirements for the higher Seismic Design Categories must be used. 
8.2.2 Seismic Design Categories D, E, and F. Structures assigned to these categories must be 
designed in anticipation of significant ductility demands that may be placed on the structures during 
their useful life. Therefore, structures in these categories are required to be designed to meet special 
detailing requirements as referenced in this section. 
8.4 COLD-FORMED STEEL 
The allowable stress and allowable load levels in AISI are incompatible with the force levels calculated 
in accordance with these Provisions. It is therefore necessary to modify the provisions of AISI for use 
with the Provisions. ANSI/ASCE 8 and SJI are both based on LRFD and thus are consistent with the 
force levels in the Provisions. As such, only minor modifications are needed to correlate those load 
factors for seismic loads to be consistent with the Provisions. The modifications of all of the reference 
documents affect only designs involving seismic loads. 
8.4.2 Light-frame walls. The provisions of this section apply to buildings framed with cold-formed 
steel studs and joists. Lateral resistance is typically provided by diagonal bracing (braced frames) or 
wall sheathing material. This section is only required for use in Seismic Design Categories D, E, and F. 
The required strength of connections is intended to assure that inelastic behavior will occur in the 
connected members prior to connection failure. Since pull-out of screws is a sudden or brittle type of 
failure, designs using pull-out to resist seismic loads are not permitted. Where diagonal members are 
used to resist lateral forces, the resulting uplift forces must be resolved into the foundation or other 
frame members without relying on the bending resistance of the track web. This often is accomplished 
by directly attaching the end stud(s) to the foundation, frame, or other anchorage device.
2003 Commentary, Chapter 8 
176 
Table 8.4-1 presents nominal shear values for plywood and oriented strand board attached to steel 
stud wall assemblies. Design values are determined by multiplying the nominal values by a phi (f) 
factor as presented in Sec. 8.4.2.5. These nominal values are based upon tests performed at Santa 
Clara University (Serrette, 1996). The test program included both cyclic and static tests; however, 
the values presented in Table 8.4-1 are based upon the cyclic tests as they are intended for use in 
seismic resistance. In low seismic areas where wind loads dominate, nominal values have been 
recommended for wind resistance by AISI based upon monotonic tests (Serrette, 1996). The cyclic 
tests were performed using the assemblies that static testing showed to be the most critical. The 
cyclically tested assemblies consisted of 3.5 by 1.625 in. C studs fabricated with ASTM A446 Grade 
A (33 ksi) material with a minimum base metal thickness of 0.033 in. Since the tests were 
conducted, ASTM A446 Grade A has been redesignated ASTM A653 SQ Grade 33. The test panels 
were 4 ft wide and 8 ft high, the sheathing material was applied vertically to only a single side of the 
studs, and there was no sheathing or bracing applied to the other side. 
The cyclic tests were performed using a sequential phase displacement protocol under development 
at the time of the test by an ad hoc Committee of the Structural Engineers Association of Southern 
California. Nominal values were conservatively established by taking the lowest load in the last set 
of stable hysteretic loops. It is expected that subsequent testing of steel stud shear wall assemblies 
will reduce or modify some of the restrictive limits currently proposed for the use of the system such 
as the nominal maximum thickness of the studs of 0.043 in., the maximum aspect ratio of 2:1, and 
the ability to use sheathing on both sides of the wall. 
8.4.4 Steel deck diaphragms. Since the design values for steel deck are based on allowable loads, 
it is necessary to present a method of deriving design strengths. Two f values are presented: 0.60 
for steel deck that is mechanically attached and 0.50 for welded steel deck. These factors are 
consistent with current proposals being circulated for inclusion in updates of ANSI/ASCE 8. 
8.5 STEEL CABLES 
The allowable stress levels of steel cable structures specified in ASCE 19 are modified for seismic 
load effects. 
8.6 RECOMMENDED PROVISIONS FOR BUCKLING-RESTRAINED BRACED 
FRAMES 
8.6.3 Commentary on Buckling-Restrained braced Frames (BRBF) 
8.6.3.1. Scope. Buckling-restrained braced frames are a special class of concentrically braced 
frames. Just as in Special Concentrically Braced Frames (SCBF), the centerlines of BRBF members 
that meet at a joint intersect at a point to form a complete vertical truss system that resists lateral 
forces. BRBF have more ductility and energy absorption than SCBFs because overall brace 
buckling, and its associated strength degradation, is precluded at forces and deformations 
corresponding to the design story drift. See AISC Sections 13 and 14 for the effects of buckling in 
SCBF. AISC Seismic Figure C13.1 shows possible BRBF bracing configurations; note that neither 
x-bracing nor k-bracing is an option for BRBF. 
BRBF are characterized by the ability of bracing elements to yield inelastically in compression as 
well as in tension. In BRBF the bracing elements dissipate energy through stable tensioncompression 
yield cycles (Clark et. al, 1999). Figure C8.6.3.2 shows the characteristic hysteretic 
behavior for this type of brace as compared to that of a buckling brace. This behavior is achieved 
through limiting buckling of the steel core within the bracing elements. Axial stress is de-coupled 
from flexural buckling resistance; axial load is confined to the steel core while the buckling 
restraining mechanism, typically a casing, resists overall brace buckling and restrains high-mode 
steel core buckling (rippling). 
Buckling-restrained braced frames are composed of columns, beams, and bracing elements, all of 
which are subjected primarily to axial forces. Braces of BRBF are composed of a steel core and a 
Steel Structure Design Requirements 
177 
buckling-restraining system encasing the steel core. Figure C8.6.3.1 shows a schematic of BRBF 
bracing element (adapted from Tremblay et al., 1999). More examples of BRBF bracing elements 
are found in Watanabe et al., 1988; Wada et. al., 1994; and Clark et al., 1999. The steel core within 
the bracing element is intended to be the primary source of energy dissipation. During a moderate to 
severe earthquake the steel core is expected to undergo significant inelastic deformations. 
BRBF can provide elastic stiffness that is comparable to that of EBF or SCBF. Full-scale laboratory 
tests indicate that properly designed and detailed bracing elements of BRBF exhibit symmetrical and 
stable hysteretic behavior under tensile and compressive forces through significant inelastic 
deformations (Watanabe et. al, 1988; Wada et. al, 1998; Clark et. al, 1999; Tremblay et. al, 1999). 
The ductility and energy dissipation capability of BRBF is expected to be comparable to that of 
SMF and greater than that of SCBF. This high ductility is attained by limiting buckling of the steel 
core. 
The axial yield strength of the core, Pysc, can be defined without dependence on other variables. This 
ability to control Pysc significantly reduces the adverse effects of relying on nominal yield strength 
values. Careful proportioning of braces throughout the building height can result in specification of 
required Pysc values that meet all of the strength and drift requirements of the applicable building 
code. 
These provisions are based on the use of brace designs qualified by testing. They are intended to 
ensure that braces are used only within their proven range of deformation capacity, and that yield 
and failure modes other than stable brace yielding are precluded at the maximum inelastic drifts 
corresponding to the design earthquake. For analyses performed using linear methods, the 
maximum inelastic drifts for this system are defined as those corresponding to 150 percent of the 
design story drift. For nonlinear time-history analyses, the maximum inelastic drifts can be taken 
directly from the analyses results. This approach is consistent with the linear analysis equations for 
design story drift in the 1997 Uniform Building Code and the 2003 NEHRP Recommended 
Provisions. It is also noted that the consequences of loss of connection stability due to the actual 
seismic displacements exceeding the calculated values may be severe; braces are therefore required 
to have a larger deformation capacity than directly indicated by linear static analysis. 
These provisions have been written assuming that future editions of NEHRP Recommended 
Provisions and of national codes will define system coefficients and limits for BRBFs. The assumed 
values for the response modification coefficient, system over strength factor, are deflection 
amplification factor are 8, 2, and 5.5 respectively. Height limits matching those for eccentrically 
braced frames are also expected. 
The design engineer utilizing these provisions is strongly encouraged to consider the effects of 
configuration and proportioning of braces on the potential formation of building yield mechanisms. 
It is also recommended that engineers refer to the following documents to gain further understanding 
of this system: Watanabe et al., 1988, Reina et al., 1997, Clark et al., 1999, Tremblay et al., 1999, 
and Kalyanaraman et al., 1998. 
During the planning stages of either a subassemblage or uniaxial brace test, certain conditions may 
exist that cause the test specimen to deviate from the parameters established in the testing appendix. 
These conditions may include: 
 Availability of beam, column, and brace sizes that reasonably match those to be used in the 
actual building frame 
 Test set-up limitations in the laboratory 
 Actuator and reaction-block capacity of the laboratory 
 Transportation and field-erection constraints 
 Actuator to subassemblage connection conditions that require reinforcement of test 
2003 Commentary, Chapter 8 
178 
specimen elements not reinforced in the actual building frame 
In certain cases, both building official and qualified peer reviewer may deem such deviations 
acceptable. The cases in which such deviations are acceptable are project-specific by nature and, 
therefore, do not lend themselves to further description in this Commentary. For these specific 
cases, it is recommended that the engineer of record demonstrate that the following objectives are 
met: 
 Reasonable relationship of scale 
 Similar design methodology 
 Adequate system strength 
 Stable buckling-restraint of the steel core 
 Adequate rotation capacity 
 Adequate cumulative strain capacity 
8.6.3.2. Bracing Members 
8.6.3.2.1 Composition 
8.6.3.2.1.1 Steel Core. The steel core is composed of a yielding segment and steel core projections; 
it may also contain transition segments between the projections and yielding segment. The area of 
the yielding segment of the steel core is expected to be sized so that its yield strength is fairly close 
to the demand calculated from the applicable building code base shear. Designing braces close to 
the predicted required strengths will help ensure distribution of yielding over multiple stories in the 
building. Conversely, over-designing some braces more than others (e.g., by using the same size 
brace on all floors), may result in an undesirable concentration of inelastic deformations in only a 
few stories. The length and area of the yielding segment, in conjunction with the lengths and areas 
of the non-yielding segments, determine the stiffness of the brace. The yielding segment length and 
brace inclination also determines the strain demand corresponding to the design story drift. 
In typical brace designs, a projection of the steel core beyond its casing is necessary in order to 
accomplish a connection to the frame. Buckling of this unrestrained zone is an undesirable yield 
mode and must therefore be precluded. 
In typical practice, the designer specifies the core plate dimensions as well as the steel material and 
grade. The steel stress-strain characteristics may vary significantly within the range permitted by the 
steel specification, potentially resulting in significant brace overstrength. This overstrength must be 
addressed in the design of connections as well as of frame beams and columns. 
In order to reduce this source of overstrength, the designer may choose to specify a brace capacity 
corresponding to a defined displacement (typically 150 percent of the design story drift) in the 
design documents. In addition, the designer may specify a limited range of acceptable yield stress in 
order to more strictly define the permissible range of core plate area. The brace supplier may then 
select the final core plate dimensions to meet the capacity requirement using the mill certificate or 
the results of a coupon test. The designer should be aware that this approach may result in a 
deviation from the calculated brace axial stiffness. The maximum magnitude of the deviation is 
dependant on the range of acceptable material yield stress. Designers following this approach should 
consider the possible range of stiffness in the building analysis in order to adequately address both 
the building period and expected drift. 
8.6.3.2.1.2 Buckling-Restraining System. This term describes those elements providing brace 
stability against overall buckling. This includes the casing as well as elements connecting the core. 
The adequacy of the buckling-restraining system must be demonstrated by testing. 
8.6.3.2.2 Testing. Testing of braces is considered necessary for this system. The applicability of 
Steel Structure Design Requirements 
179 
tests to the designed brace is defined in Sec. 8.6.3.7. Sec. 9.2a, which describes in general terms the 
applicability of tests to designs, applies to BRBF. 
BRBF designs require reference to successful tests of a similarly-sized test specimen and of a brace 
subassemblage that includes rotational demands. The former is a uniaxial test intended to 
demonstrate adequate brace hysteretic behavior. The latter is intended to verify the general brace 
design concept and demonstrate that the rotations associated with frame deformations do not cause 
failure of the steel core projection, binding of the steel core to the casing, or otherwise compromise 
the brace hysteretic behavior. A single test may qualify as both a subassemblage and a brace test 
subject to the requirements of Sec. 8.6.3.7; for certain frame-type subassemblage tests, obtaining 
brace axial forces may prove difficult and separate brace tests may be necessary. A sample 
subassemblage test is shown in Figure C8.6.3.5 (from Tremblay, 1999). 
Tests cited serve another function in the design of BRBF: the maximum forces that the brace can 
deliver to the system are determined from test results. Calculation of these maximum forces is 
necessary for connection design and for the design of beams in V- and inverted-V configurations (see 
Sec. 8.6.3.4.1.3). In order to permit a realistic design of these beams, two separate calculations are 
made. The compression-strength adjustment factor, , accounts for the compression overstrength 
(with respect to tension strength) noted in buckling-restrained braces in recent testing (SIE, 1999). 
The tension strength adjustment factor, ., accounts for material overstrength (Ry) and strain 
hardening. Figure C-8.3.6.3 shows a diagrammatic bilinear force-displacement relationship in 
which the compression strength adjustment factor  and the tension-strength adjustment factor . are 
related to brace forces and nominal material yield strength. These quantities are defined as: 
max 
max 
y 
y 
F A P 
F A T 
. 
 
. 
= = 
y max 
y y 
F A T 
FA FA 
.=. = 
where Pmax is the maximum compression force and Tmax is the maximum tension force within 
deformations corresponding to 150 percent of the Design Story Drift (these deformations are defined 
as 1.5Dbm in the Appendix on testing). The acceptance criteria for testing require that values of  and 
. be greater than or equal to 1.0 for buckling-restrained braces. 
8.6.3.2.3 Quality Assurance. The design provisions for BRBFs are predicated on reliable brace 
performance. In order to assure this performance, a quality assurance plan is required. These 
measures are in addition to those covered in the code of standard practice and Sec. 18. Examples of 
measures that may provide quality assurance are: 
 Special inspection of brace fabrication. 
 Inspection may include confirmation of fabrication and alignment tolerances, as well as 
NDT methods for evaluation of the final product. 
 Brace manufacturers participation in a recognized quality certification program. 
 Certification should include documentation that the manufacturers quality assurance plan is 
in compliance with the requirements of the BRBF Provisions, the Seismic Provisions for 
Structural Steel Buildings, and the Code of Standard Practice. The manufacturing and 
quality control procedures should be equal to, or better, than those used to manufacture 
brace test specimens. 
2003 Commentary, Chapter 8 
180 
8.6.3.3 Bracing Connections. Bracing connections must not yield at force levels corresponding to 
the yielding of the steel core; they are therefore designed for the maximum force that can be 
expected from the brace. In the actual building frame, the use of slip-critical bolts designed at 
factored loads is encouraged (but not required) to greatly reduce the contribution of bolt slip to the 
total inelastic deformation in the brace. Because of the way bolt capacities are calibrated, the 
engineer should recognize that the bolts are going to slip at load demands 30 percent lower than 
published factored capacities. This slippage is not considered to be detrimental to behavior of the 
BRBF system and is consistent with the design approach found elsewhere in Sec. 7.2 of AISC 
Seismic. See also commentary on Sec. C7.2 of AISC Seismic. Bolt holes may be drilled or punched 
subject to the requirements of LRFD Specification Sec. M2.5. 
8.6.3.4 Special Requirements Related to Bracing Configuration 
8.6.3.4.1. In SCBF, V-bracing has been characterized by a change in deformation mode after one of 
the braces buckles. This is due to the negative post-buckling stiffness, as well as the difference 
between tension and compression capacity, of traditional braces. Since buckling-restrained braces 
do not exhibit the negative secant stiffness associated with post-buckling deformation, and have only 
a small difference between tension and compression capacity, the practical requirements of the 
design provisions for this configuration are relatively minor. Figure C8.6.3.4 shows the deformation 
mode that develops after one brace has yielded but before the yielding of the opposite brace 
completes the mechanism. This mode involves flexure of the beam and elastic axial deformation of 
the un-yielded brace; it also involves inelastic deformation of the yielded brace that is much greater 
than the elastic deformation of the opposing brace. The drift range that corresponds to this 
deformation mode depends on the flexural stiffness of the beam. Therefore, where V-braced frames 
are used, it is required that a beam be provided that has sufficient stiffness, as well as strength, to 
permit the yielding of both braces within a reasonable story drift considering the difference in 
tension and compression capacities determined by testing. 
The beam is expected to undergo this deflection, which is permanent, during moderate seismic 
events; a limit is therefore applied to this deflection. Additionally, the required brace deformation 
capacity must include the additional deformation due to beam deflection under this load. Since other 
requirements such as the brace testing protocol (Sec. 8.6.3.7.6.3) and the stability of connections 
(Sec.8.6.3.3.3) depend on this deformation, engineers will find significant incentive to avoid flexible 
beams in this configuration. Where the special configurations shown in Figure C13.3 of AISC 
Seismic are used, the requirements of this section are not relevant. 
8.6.3.5 Columns. Columns in BRBF are required to have compact sections because some inelastic 
rotation demands are possible. Columns are also required to be designed considering the maximum 
force that the adjoining braces are expected to develop. 
8.6.3.6 Beams. Like columns, beams in BRBF are required to have compact sections because some 
inelastic rotation demands are possible. Likewise, they are also required to be designed considering 
the maximum force that the adjoining braces are expected to develop. 
8.6.3.7.1 Scope and Purpose. Development of the testing requirements in these provisions was 
motivated by the relatively small amount of test data on this system available to structural engineers. 
In addition, no data from the response of BRBFs to severe ground motion is available. Therefore, 
the seismic performance of these systems is relatively unknown compared to more conventional 
steel-framed structures. 
The behavior of a buckling restrained brace frame differs markedly from conventional braced frames 
and other structural steel seismic-force-resisting systems. Various factors affecting brace 
performance under earthquake loading are not well understood and the requirement for testing is 
intended to provide assurance that the braces will perform as required, and also to enhance the 
overall state of knowledge of these systems. 
Steel Structure Design Requirements 
181 
It is recognized that testing of brace specimens and subassemblages can be costly and timeconsuming. 
Consequently, this Chapter has been written with the simplest testing requirements 
possible, while still providing reasonable assurance that prototype BRBFs based on brace specimens 
and subassemblages tested in accordance with these provisions will perform satisfactorily in an 
actual earthquake. 
It is not intended that these provisions drive project-specific tests on a routine basis for building 
construction projects. In most cases, tests reported in the literature, or supplied by the brace 
manufacturer, can be used to demonstrate that a brace and subassemblage configuration satisfies the 
strength and inelastic rotation requirements of these provisions. Such tests, however, should satisfy 
the requirements of this Chapter. 
The provisions have been written allowing submission of data on previously tested, based on 
similarity conditions. As the body of test data for each brace type grows, the need for additional 
testing is expected to diminish. The provisions allow for manufacturer-designed braces, through the 
use of the design methodology. 
Most testing programs developed for primarily axial-load-carrying components focus largely on 
uniaxial testing. However, these provisions are intended to direct the primary focus of the program 
toward testing of a subassemblage that imposes combined axial and rotational deformations on the 
brace specimen. This reflects the view that the ability of the brace to accommodate the necessary 
rotational deformations cannot be reliably predicted by analytical means alone. Subassemblage test 
requirements are discussed more completely in Sec. 8.6.3.7.4. 
Where conditions in the actual building differ significantly from the test conditions specified in this 
Chapter, additional testing beyond the requirements described herein may be needed to assure 
satisfactory brace performance. Prior to developing a test program, the appropriate regulatory 
agencies should be consulted to assure the test program meets all applicable requirements. 
8.6.3.7.2 Symbols. The provisions require the introduction of several new variables. The quantity 
.bm represents both an axial displacement and a rotational quantity. Both quantities are determined 
by examining the profile of the building at the design story drift, .m, and extracting joint lateral and 
rotational deformation demands. 
Determining the maximum rotation imposed on the braces used in the building may require 
significant effort. The engineer may prefer to select a reasonable value (i.e., interstory drift), which 
can be simply demonstrated to be conservative for each brace type, and is expected to be within the 
performance envelope of the braces selected for use on the project. 
The brace deformation at first significant yield is used in developing the test sequence described in 
Sec. 8.6.3.7.6.3. The quantity is required to determine the actual cumulative inelastic deformation 
demands on the brace. If the nominal yield stress of the steel core were used to determine the test 
sequence, and significant material over-strength were to exist, the total inelastic deformation 
demand imposed during the test sequence would be overestimated. 
8.6.3.7.3 Definitions. Two types of testing are referred to in this Chapter. The first type is 
subassemblage testing, described in Sec. 8.6.3.7.4, an example of which is illustrated in Figure 
C8.6.3.5. 
The second type of testing described in Sec. 8.6.3.7.5 as brace specimen testing is permitted to be 
uniaxial testing. 
8.6.3.7.4 Subassemblage Test Specimen. The objective of subassemblage testing is to verify the 
ability of the brace, and in particular its steel core extension and buckling restraining mechanism, to 
accommodate the combined axial and rotational deformation demands without failure. 
It is recognized that subassemblage testing is more difficult and expensive than uniaxial testing of 
brace specimens. However, the complexity of the brace behavior due to the combined rotational and 
2003 Commentary, Chapter 8 
182 
axial demands, and the relative lack of test data on the performance of these systems, indicates that 
subassemblage testing should be performed. 
Subassemblage testing is not intended to be required for each project. Rather, it is expected that 
brace manufacturers will perform the tests for a reasonable range of axial loads, steel core 
configurations, and other parameters as required by the provisions. It is expected that this data will 
subsequently be available to engineers on other projects. Manufacturers are therefore encouraged to 
conduct tests that establish the device performance limits to minimize the need for subassemblage 
testing on projects. 
Similarity requirements are given in terms of measured axial yield strength of both the prototype and 
the test specimen braces. This is better suited to manufacturers product testing than to projectspecific 
testing. Comparison of mill certificate or coupon test results is a way to establish a 
similarity between the Subassemblage Test Specimen brace and the Prototype braces. Once 
similarity is established, it is acceptable to fabricate Test Specimens and Prototype braces from 
different heats of steel. 
A variety of subassemblage configurations are possible for imposing combined axial and rotational 
deformation demands on a test specimen. Some potential subassemblages are shown in Figure 
C8.3.6. The subassemblage need not include connecting beams and columns provided that the test 
apparatus duplicates, to a reasonable degree, the combined axial and rotational deformations 
expected at each end of the brace. 
Rotational demands may be concentrated in the steel core extension in the region just outside the 
buckling restraining mechanism. Depending on the magnitude of the rotational demands, limited 
flexural yielding of the steel core extension may occur. Rotational demands can also be 
accommodated by other means, such as tolerance in the buckling restraint layer or mechanism, 
elastic flexibility of the brace and steel core extension, or through the use of pins or spherical 
bearing assemblies. It is in the engineers best interest to include in a subassemblage testing all 
components that contribute significantly to accommodating rotational demands. The use of pins, 
while accommodating rotational demands, creates the potential for instability; and should be 
carefully considered by the engineer. 
It is intended that the subassemblage test specimen be larger in axial-force capacity than the 
Prototype. However, the possibility exists for braces to be designed with very large axial forces. 
Should the brace yield force be so large as to make subassemblage testing impractical, the engineer 
is expected to make use of the provisions that allow for alternate testing programs, based on building 
official approval and qualified peer review. Such programs may include, but are not limited to, nonlinear 
finite element analysis, partial specimen testing, and reduced-scale testing, in combination 
with full-scale uniaxial testing where applicable or required. 
The steel core material was not included in the list of requirements. The more critical parameter, 
calculated margin of safety for the steel core projection stability, is required to meet or exceed the 
value used in the prototype. The method of calculating the steel core projection stability should be 
included in the design methodology. 
8.6.3.7.5 Brace Test Specimen. The objective of brace test specimen testing is to establish basic 
design parameters for the BRBF system. 
It is recognized that the fabrication tolerances used by brace manufacturers to achieve the required 
brace performance may be tighter than those used for other fabricated structural steel members. The 
engineer is cautioned against including excessively prescriptive brace specifications, as the intent of 
these provisions is that the fabrication and supply of the braces is achieved through a performancebased 
specification process. It is considered sufficient that the manufacture of the test specimen and 
the prototype braces be conducted using the same quality control and assurance procedures, and the 
braces be designed using the same design methodology. 
Steel Structure Design Requirements 
183 
The engineer should also recognize that manufacturer process improvements over time may result in 
some manufacturing and quality control and assurance procedures changing between the time of 
manufacture of the brace test specimen and of the prototype. In such cases reasonable judgment is 
required. 
If the steel core or steel core projection is not biaxially symmetric, the engineer should ensure that 
the same orientation is maintained in both the test specimen and the prototype. 
The allowance of previous test data (similarity) to satisfy these provisions is less restrictive for 
uniaxial testing than for subassemblage testing. Subassemblage test specimen requirements are 
described in Sec. C8.6.3.7.4. 
A considerable number of uniaxial tests have been performed on some brace systems and the 
engineer is encouraged, wherever possible, to submit previous test data to meet these provisions. 
Relatively few Subassemblage tests have been performed. This type of testing is considered a more 
demanding test of the overall brace performance. 
8.6.3.7.5.4 Connection Details. In many cases it will not be practical or reasonable to test the exact 
brace connections present in the prototype. These provisions are not intended to require such 
testing. In general, the demands on the steel core extension to gusset-plate connection are well 
defined due to the known axial capacity of the brace and the limited flexural capacity of the steel 
core extension. The subsequent design of the bolted or welded connection is relatively wellunderstood 
and it is not intended that these connections become the focus of the testing program. 
For the purposes of utilizing previous test data to meet the requirements of this Chapter, the 
requirements for similarity between the brace and subassemblage brace test specimen can be 
considered to exclude the steel core extension connection to frame. 
8.6.3.7.5.5 Materials. The intent of the provisions is to allow test data from previous test programs 
to be presented where possible. See Sec.8.6.3.7.4 for additional commentary. 
8.6.3.7.5.7 Bolts. For the brace test specimen, it is crucial to treat the ultimate load that can be 
expected in the braces as the load at which bolt slippage should be prevented. Prevention of bolt 
slippage increases the chances of achieving a successful test and protects laboratory setup. In terms 
of the nomenclature used by the Research Council on Structural Connections (RCSC), prevention of 
bolt slippage implies using service-level load capacities when sizing bolted connections. Bolted 
connections sized using service-level capacities per RCSC will provide at least a 90 percent 
reliability that the bolts will not slip at the maximum force developed by the braces during the test. 
The intent of this provision is to ensure that the bolted end-connections of the brace test specimen 
reasonably represent those of the prototype. It is possible that due to fabrication or assembly 
constraints variations in faying-surface preparation, bolt-hole fabrication, and bolt size may occur. In 
certain cases, such variations may not be detrimental to the qualification of a successful cyclic test. 
Final acceptability of variations in brace-end bolted connection rest on the opinion of the building 
official or qualified peer reviewer. 
8.6.3.7.6.3 Loading Sequence. The subassemblage test specimen is required to undergo combined 
axial and rotational deformations similar to those in the prototype. It is recognized that identical 
braces, in different locations in the building, will undergo different maximum axial and rotational 
deformation demands. In addition, the maximum rotational and axial deformation demands may be 
different at each end of the brace. The engineer is expected to make simplifying assumptions to 
determine the most appropriate combination of rotational and axial deformation demands for the 
testing program. 
Some subassemblage configurations will require that one deformation quantity be fixed while the 
other is varied as described in the test sequence above. In such a case, the rotational quantity may be 
applied and maintained at the maximum value, and the axial deformation applied according to the 
2003 Commentary, Chapter 8 
184 
test sequence. The engineer may wish to perform subsequent tests on the same subassemblage 
specimen to bound the brace performance. 
The loading sequence requires each tested brace to achieve ductilities corresponding to 1.5 times the 
design story drift and a cumulative inelastic axial ductility capacity of 140. Both of these 
requirements are based on a study in which a series nonlinear dynamic analyses was conducted on 
model buildings in order to investigate the performance of this system; the ductility capacity 
requirement represents a mean of response values and the cumulative ductility capacity requirement 
is a mean plus standard deviation value (Sabelli, 2001). In that study, buildings were designed and 
models of brace hysteresis selected so as to maximize the demands on braces. It is therefore 
believed that these requirements are more severe than the demands that typical braces in typical 
designs would face under their design-basis ground motion, perhaps substantially so. It is also 
expected that as more test data and building analysis results become available these requirements 
may be revisited. 
The ratio of brace yield deformation (Dby) to the brace deformation corresponding to the design story 
drift (Dbm) must be calculated in order to define the testing protocol. This ratio is typically the same 
as the ratio of the displacement amplification factor (as defined in the applicable building code) to 
the actual overstrength of the brace; the minimum overstrength is defined in Sec. 8.6.3.2.1.1.1. 
Engineers should note that there is a minimum brace deformation demand corresponding to 1% 
story drift (8.6.3.7.2); provision of overstrength beyond that required to so limit the design story drift 
may not be used as a basis to reduce the testing protocol requirements. 
Table C8.6.3.7.1 shows an example brace test protocol. For this example, it is assumed that the 
brace deformation corresponding to the design story drift is four times the yield deformation; it is 
also assumed that the design story drift is larger than the 1 percent minimum. The test protocol is 
then constructed from steps 1-4 of Sec. 8.6.3.7.6.3. In order to calculate the cumulative inelastic 
deformation, the cycles are converted from multiples of brace deformation at the design story drift 
(Dbm) to multiples of brace yield deformation (Dby). Since the cumulative inelastic drift at the end of 
the 1.5Dbm cycles is less than the minimum of 140Dby required for brace tests, additional cycles to 
Dbm are required. At the end of three such cycles, the required cumulative inelastic deformation has 
been reached. 
Table C8.6.3.7.1 Example Brace Testing Protocol 
Cycle 
Deformation 
Deformation 
Inelastic 
Deformation 
Inelastic Cumulative 
6 @ Dby= 0Dby = 6*4*(Dby  Dby) = 0Dby 0Dby 
4 @ 0.5Dbm= 16Dby = 4 @ 2.0Dby = 4*4*(2.0Dby  Dby) = 16Dby 0Dby +16Dby 
4 @ Dbm= 64Dby = 4 @ 4.0Dby = 4*4*(4.0Dby  Dby) = 48Dby 16Dby +48Dby 
2 @ 1.5Dbm= 104Dby = 2 @ 6.0Dby = 2*4*(6.0Dby  Dby) = 40Dby 64Dby +40Dby 
3 @ Dbm= 140Dby = 3 @ 4.0Dby = 3*4*(4.0Dby  Dby) = 36Dby 104Dby +36Dby 
Cumulative Inelastic Deformation at End of Protocol = 140 Dby 
Dynamically applied loads are not required by these provisions. The use of slowly applied cyclic 
loads, widely described in the literature for brace specimen tests, is acceptable for the purposes of 
these provisions. It is recognized that dynamic loading can considerably increase the cost of testing, 
and that few laboratory facilities have the capability to apply dynamic loads to very large-scale test 
specimens. Furthermore, the available research on dynamic loading effects on steel test specimens 
has not demonstrated a compelling need for such testing. 
If rate-of-loading effects are thought to be potentially significant for the steel core material used in 
the prototype, it may be possible to estimate the expected change in behavior by performing coupon 
tests at low (test cyclic loads) and high (dynamic earthquake) load rates. The results from brace 
tests would then be factored accordingly. 
Steel Structure Design Requirements 
185 
8.6.3.7.8 Materials Testing Requirements. Tension testing of the steel core material used in the 
manufacture of the test specimens is required. In general, there has been good agreement between 
coupon test results and observed tensile yield strengths in full-scale uniaxial tests. Material testing 
required by this appendix is consistent with that required for testing of beam-to-column moment 
connections. For further information on this topic refer to Section CS8 of AISC Seismic. 
8.6.3.7.10 Acceptance Criteria. The acceptance criteria are written so that the minimum testing 
data that must be submitted is at least one subassemblage test and at least one uniaxial test. In most 
cases the subassemblage test also qualifies as a uniaxial test provided the requirements of Sec. 
8.6.3.7.5 are met. If project specific subassemblage testing is to be performed it may be simplest to 
perform two subassemblage tests to meet the requirements of this section. For the purposes of these 
requirements a single subassemblage test incorporating two braces in a chevron or other 
configuration is also considered acceptable. 
Depending on the means used to connect the test specimen to the subassemblage or test apparatus, 
and the instrumentation system used, bolt slip may appear in the load vs. displacement history for 
some tests. This may appear as a series of spikes in the load vs. displacement plot and is not 
generally a cause for concern, provided the behavior does not adversely affect the performance of 
the brace or brace connection. 
These acceptance criteria are intended to be minimum requirements. The 1.3 limit in Sec. 
8.6.3.7.10.5 is essentially a limitation on . These provisions were developed assuming that  < 1.3 
so this provision has been included in the test requirements. Most currently available braces should 
be able to satisfy this requirement. 
Figure C8.6.3.1 Detail of a Buckling Restrained Brace (Courtesy of R. Tremblay). 
PLATE 
BOND-PREVENTING 
LAYER 
PLATE 
PLATE 
TRANSITION SEGMENT YIELDING SEGMENT 
BOND-PREVENTING 
LAYER 
PLATE 
CORE PROJECTION
2003 Commentary, Chapter 8 
186 
Figure 8.6.3.2 Detail of buckling-restrained (unbonded) brace hysteretic behavior 
(Courtesy of Seismic Isolation Engineering). 
Figure C8.6.33 Diagram of brace force-displacement. 
Compression Tension 
-Dbm 
Tmax= wFyA 
Ty = RyFyA 
Py 
Pmax=-bwFyA 
-Dby Dby Dbm
2003 Commentary, Chapter 8 
187 
Figure C8.6.3.4 Post-yield, pre-mechanism change in deformation mode 
for V- and Inverted-V BRBF. 
Figure C8.6.3.5 Example of test subassemblage. 
PINNED JOINT 
(typ.) 
900 kN ACTUATOR 
(typ.) 
BRACING 
(typ.) 
ANCHOR BLOCK 
(typ.) 
PINNED JOINT 
(typ.)
2003 Commentary, Chapter 8 
188 
Figure C8.6.3.6 Schematic of possible test subassemblages. 
Eccentric Loading 
of Brace 
Loading of Brace 
with 
Constant Imposed 
Loading of Braced 
Frame 
Loading of Brace and 
Column
Steel Structure Design Requirements 
189 
8.7. SPECIAL STEEL PLATE WALLS 
8.7.3 Scope. These provisions for SSPWs are intended for use in conjunction with AISC 
Seismic. 
In special steel plate walls (SSPWs), the slender unstiffened steel plates (webs) connected to 
surrounding horizontal and vertical boundary elements (i.e., HBEs and VBEs) are designed to 
yield and behave in a ductile hysteretic manner during earthquakes. All HBEs are also rigidly 
connected to the VBEs with moment resisting connections able to develop the expected plastic 
moment of the HBEs. Each web must be surrounded by boundary elements. 
Figure C8.7.3.1 Schematic of Special Steel Plate Walls. 
Experimental research on SSPWs subjected to cyclic inelastic quasi-static and dynamic loading 
(Thorburn, et al., 1983; Timler and Kulak, 1983; Tromposch and Kulak, 1987; Roberts and 
Sabouri-Ghomi, 1992; Cassese et al., 1993; Driver et al., 1997; Elgaaly 1998; Rezai, 1999; 
Lubell et al., 2000;) has demonstrated their ability to behave in a ductile manner and dissipate 
significant amounts of energy. This has been confirmed by analytical studies using finite 
element analysis and other analysis techniques (Sabouri-Ghomi and Roberts, 1992; Elgaaly et 
al., 1993; Elgaaly and Liu, 1997; Driver et al., 1997). 
Yielding of the webs occurs by development of tension field action at an angle close to 45 
from the vertical, and buckling of the plate in the orthogonal direction. Past research shows 
that the sizing of VBEs and HBEs in a SSPW makes it possible to develop this tension field 
action across the entire webs. Except for cases with very stiff HBEs and VBEs, yielding in the 
webs develops in a progressive manner across each panel. Because the webs do not yield in 
compression, continued yielding upon repeated cycles of loading is contingent upon the SSPW 
being subjected to progressively larger drifts, except for the contribution of plastic hinging 
developing in the HBEs to the total system hysteretic energy. In past research (Driver et al. 
1997), the yielding of boundary elements contributed approximately 25-30 percent of the total 
load strength of the system. 
Level 1 
Level 2 
Level 3 
Level 4 
Horizontal Boundary 
Element (HBE) 
Vertical Boundary 
Element (VBE) 
Web Element 
Panel 1 
Panel 2 
Panel 3 
Panel 4
2003 Commentary, Chapter 8 
190 
With the exception of plastic hinging at the ends of HBEs, the surrounding horizontal and 
vertical boundary elements are designed to remain essentially elastic when the webs are fully 
yielded. Plastic hinging at the ends of HBEs is needed to develop the plastic collapse 
mechanism of this system. Plastic hinging in the middle of HBEs, which could partly prevent 
yielding of the webs, is deemed undesirable. Cases of both desirable and undesirable yielding 
in VBEs have been observed in past testing. In absence of a theoretical formulation to quantify 
the conditions leading to acceptable yielding (and supporting experimental validation of this 
formulation), the conservative requirement of elastic VBE response is justified. 
Research literature often compares the behavior of steel plate walls to that of a vertical plate 
girder, indicating that the webs of a SSPW resist shears by tension field action (similarly to the 
webs of a plate girder) and that the VBEs of a SSPW resist overturning moments (similarly to 
the flanges of a plate girder). While this analogy is useful in providing a conceptual 
understanding of the behavior of SSPWs, many significant differences exist in the behavior and 
strength of the two systems. Past research shows that the use of structural shapes for the VBEs 
and HBEs in SSPWs (as well as other dimensions and details germane to SSPWs) favorably 
impacts orientation of the angle of development of the tension field action, and makes possible 
the use of very slender webs (having negligible diagonal compressive strength). Sizeable top 
and bottom HBEs are also required in SSPWs to anchor the significant tension fields that 
develop at these ends of the structural system. Limits imposed on the maximum web 
slenderness of plate girders to prevent flange buckling, or due to transportation requirements, 
are also not applicable to SSPWs which are constructed differently. For these reasons, the use 
of Appendix G in AISC LRFD Specifications for the design of SSPW is not appropriate. 
8.7.4 Webs. The specified minimum yield stress of steel used for SSPW is per Sec. 6.1. 
However, the webs of SSPWs could also be of special highly ductile low yield steel having 
specified minimum yield in the range of 12 to 33 ksi. 
8.7.4.1. The lateral shears are carried by tension fields that develop in the webs stressing in the 
direction a, defined in Sec. 8.7.4. To determine a, when the HBEs and VBEs boundary 
elements of a web are not identical, the average of HBE areas may be taken in the calculation 
of Ab, and the average of VBE areas and inertias may be respectively used in the calculation Ac, 
and Ic. 
Plastic shear strength of panels is given by 0.5 Ry Fy tw Lcf sin2 a. Nominal strength is obtained 
by dividing this value by a system overstrength, as defined by FEMA 369, and taken as 1.2 for 
SSPWs (Berman and Bruneau 2003). 
The above plastic shear strength is obtained from the assumption that, for purposes of analysis, 
each web may be modeled by a series of inclined pin-ended strips Figure C8.7.4-1, oriented at 
angle a. Past research has shown this model to provide realistic results, as shown in Figure 
C8.7.4-2 for example, provided at least 10 equally spaced strips are used to model each panel. 
Steel Structure Design Requirements 
191 
Figure C8.7.4-1 Strip model of a SSPW. 
Figure C8.7.4-2 Comparison of Experimental Results and 
Strength predicted by Strip Model (Driver et al. 1997). 
8.7.4.2 Panel Aspect Ratio. Past research shows that modeling SSPWs with strips is 
reasonably accurate for panel aspect ratios of L/h that exceed 0.8 (Rezai 1999). Additional 
horizontal intermediate boundary elements could be introduced in SSPW to modify the L/h of 
panels having such an aspect ratio less than 0.8. 
h 
L 
a 
DRIVER ET. AL.'S TEST RESULTS FOR 
LOWER PANEL OF MULTI-STOREY SPSW FRAME 
-4000 
-3000 
-2000 
1000
0 
-1000 
2000 
3000 
4000
-60 -40 -20 0 20 40 60 80 
Test 
Strip Model: Assembly 
Strip Model: Frame Only 
Panel Deformation (mm) 
Panel Shear (kN)
2003 Commentary, Chapter 8 
192 
No theoretical upper bound exists on L/h (provided sufficiently stiff HBEs can be provided), 
but a maximum value of 2.5 is specified on the basis that past research has not investigated the 
seismic behavior of SSPWs having L/h greater than 2.0. Excessive flexibility of HBEs is of 
concern for L/h ratio beyond the specified limit. For conditions beyond the specified limits, 
other FEM methods shall be used which correlate with published test data. 
8.7.4.2 Openings in webs. Large openings in webs create significant local demands and thus 
must have HBE and VBE in a similar fashion as the remainder of the system. When openings 
are required, SSPWs can be subdivided in smaller SSPW segments by using HBEs and VBEs 
bordering the openings. SSPWs with holes in the web not surrounded by HBEs/VBEs have not 
been tested. The provisions will allow other openings that can be justified by analysis or 
testing. 
8.7.4.4 Maximum Slenderness Ratio for Plates. A limit on the slenderness of plate elements 
in SSPW is needed to ensure that adequate cyclic ductility is provided. The limit provided is 
consistent with successful tests of this system. 
8.7.5 Connections of Webs to Boundary Elements. The required strength of web 
connections to the surrounding HBEs and VBEs shall develop the expected tensile strength of 
the webs. Net-sections shall also provide this strength for the case of bolted connections. 
The strip model can be used to model the behavior of SSPWs and the tensile yielding of the 
webs at angle a. A single angle of inclination taken as the average for all the panels may be 
used to analyze the entire wall. The expected tensile strength of the web strips shall be defined 
as 
RyFyAs 
where As is the area of a strip, equal to 
As = (L cos a + H sin a) / n 
where L and H are the width and height of a panel, n is the number of strips per panel, and n 
shall be taken greater than or equal to 10. 
This analysis method has been shown, though correlation with physical test data, to adequately 
predict SSPW performance. It is recognized, however, that other advanced analytical 
techniques (such as the Finite Element Method (FEM)) may also be used for design of SSPWs. 
If such non-linear (geometric and material) FEM models are used, they should be calibrated 
against published test results to ascertain reliability for application. 
8.7.6 Horizontal and Vertical Boundary Elements (HBEs and VBEs). Per capacity design 
principles, all edge boundary elements (HBEs and VBEs) shall be designed to resist the 
maximum forces generated by the tension field action of the webs fully yielding. Axial forces, 
shears, and moments develop in the boundary elements of the SSPW as a result of the response 
of the system to the overall overturning and shear, and this tension field action in the webs. 
Actual web thickness must be considered for this calculation, because webs thicker than 
required may have be used due to availability, or minimum thickness required for welding. 
At the top panel of the wall, the vertical components of the tension field shall be anchored to 
the HBE. The HBE shall have sufficient strength to allow development of full tensile yielding 
across the panel width. 
At the bottom panel of the wall, the vertical components of the tension field shall be anchored 
to the HBE. The HBE shall have sufficient strength to allow development of full tensile 
yielding across the panel width. This may be accomplished by continuously anchoring the 
HBE to the foundation. 
Steel Structure Design Requirements 
193 
For intermediate HBEs of the wall, the anticipated variation between the top and bottom web 
normal stresses acting on the HBE is usually small, or null when webs in the panel above and 
below the HBE have identical thickness. While top and bottom HBEs are typically of 
substantial size, intermediate HBEs are relatively smaller. 
Beyond the exception mentioned in Sec. 8.7, in some instances, the engineer may be able to 
justify yielding of the boundary elements by demonstrating that the yielding of this edge 
boundary element will not cause reduction on the SSPW shear capacity to support the demand 
and will not cause a failure in vertical gravity carrying capacity. 
Forces and moments in the members (and connections), including those resulting from tension 
field action, may be determined from a plane frame analysis. The web is represented by a 
series of inclined pin-ended strips, as described in C8.7.4-1. A minimum of 10 equally spaced 
pin-ended strips per panel will be used in such an analysis. 
A number of analytical approaches are possible to achieve capacity design and determine the 
same forces acting on the vertical boundary elements. Some example methods applicable to 
SSPWs follow. In all cases, actual web thickness must be considered, for reasons described 
earlier. 
Non-linear push-over analysis 
A model of the SSPW can be constructed in which bi-linear elasto-plastic web elements of 
strength RyFyAs are introduced in the direction a. Bi-linear plastic hinges can also be 
introduced at the ends of the horizontal boundary elements. Standard push-over analysis 
conducted with this model will provide axial forces, shears, and moments in the boundary 
frame when the webs develop yielding. Separate checks are required to verify that plastic 
hinges do not develop in the horizontal boundary elements, except at their ends. 
Combined linear elastic computer programs and capacity design concept 
The following four-step procedure provides reasonable estimates of forces in the boundary 
elements of SSPW systems. 
Lateral Forces: Use combined model, boundary elements and web elements, to come up with 
the demands in the web and the boundary elements based on the code required base shear. The 
web elements shall not be considered as vertical-load carrying elements. 
Gravity Load (Dead Load and Live Load): Apply gravity loads to a model with only gravity 
frames. The web elements shall not be considered as vertical-load carrying elements. 
Without any overstrength factors, design the boundary elements using the demands based on 
combination forces of the above steps 1 and 2. 
Boundary Element Capacity Design Check: Check the boundary element for the maximum 
capacity of the web elements in combination with the maximum possible axial load due to 
over-turning moment. Use the axial force obtained from step 1 above and multiply by 
overstrength factor Oo
. Apply load from web elements (Ry Fy As) in the direction of a. For this 
capacity design check use material strength reduction factor of 1.0. 
Indirect Capacity Design Approach 
The CSA-S16-02 (CSA 2002) proposes that loads in the vertical boundary members can be 
determined from the gravity loads combined with the seismic loads increased by the 
amplification factor, 
B = Ve / Vu 

2003 Commentary, Chapter 8 
194 
where 
Ve = expected shear strength, at the base of the wall, determined for the web 
thickness supplied 
= 0.5 Ry Fy tw L sin2a 
Vu = factored lateral seismic force at the base of the wall 
In determining the loads in VBEs, the amplification factor, B, need not be taken as greater than 
R. 
The VBE design axial forces shall be determined from overturning moments defined as 
follows: 
(i) the moment at the base is BMu, where Mu is the factored seismic overturning moment at the 
base of the wall corresponding to the force Vu; 
(ii) the moment BMu extends for a height H but not less than two stories from the base; and 
(iii) the moment decreases linearly above a height H to B times the overturning moment at one 
story below the top of the wall, but need not exceed R times the factored seismic overturning 
moment at the story under consideration corresponding to the force Vu. 
The local bending moments in the VBEs due to tension field action in the web shall be 
multiplied by the amplification factor B. 
Preliminary Design 
For preliminary proportioning of HBEs, VBEs, and webs, a SSPW wall may be approximated 
by a vertical truss with tension diagonals. Each web is represented by a single diagonal tension 
brace within the story. For an assumed angle of inclination of the tension field, the web 
thickness, tw, may be taken as: 
a 
. 
sin 2 
2 sin 
L
t A s 
w 
O 
= 
where 
A = area of the equivalent tension brace 
. = angle between the vertical and the longitudinal axis of the equivalent diagonal 
brace 
L = the distance between VBE centerlines 
a = assumed angle of inclination of the tension field measured from the vertical per 
Sec. 8.7.4 
Os 
= the system overstrength factor as defined by FEMA 369 and taken as 1.2 for 
SSPWs (Berman and Bruneau 2003). 
Determination of A is originally estimated from an equivalent brace size to meet the structures 
drift requirements. 
8.7.6.3 Boundary Element Compactness. Some amount of local yielding is expected in the 
HBEs and VBEs to allow development of the plastic mechanism of SSPW systems. For that 
reason, HBEs and VBEs shall comply with the requirements in AISC Seismic Table I-8-1 for 
SMFs. 
8.7.6.5 Lateral Bracing. Providing stability of SSPW systems boundary elements is necessary 
for proper performance of the system. The lateral bracing requirements for HBEs are provided 
Steel Structure Design Requirements 
195 
to be consistent with beams in SMFs for both strength and stiffness. In addition, all 
intersections of HBEs and VBEs must be braced to ensure stability of the entire panel. 
8.7.6.7 Panel Zone. Panel zone requirements are not imposed for intermediate HBEs. These 
are expected to be small HBEs connecting to sizeable VBEs. The engineer should use 
judgment to identify special situations in which the panel zone adequacy of VBEs next to 
intermediate HBEs should be verified. 
8.7.6.8 Stiffness of Vertical Boundary Elements. This requirement is intended to prevent 
excessive in-plane flexibility and buckling of VBEs. Opportunity exists for future research to 
confirm or improve the applicability this requirement. 
REFERENCES 
Astaneh, A., 1998, Seismic Behavior and Design of Gusset Plates for Braced Frames, Steel 
Tips, Structural Steel Education Council, Moraga, California. 
Astaneh-Asl, A., and Zhao, Q. (2002) "Cyclic Behavior of Steel Shear WallSystems ", 
Proceedings, 2002 Annual Stability Conference, StructuralStability Research Council, April 
25-27, Seattle. 
Astaneh-Asl, A. (2001) "Seismic Behavior and Design of Steel Shear Walls",Steel Technical 
Information and Product Services Report (Steel TIPS),Structural Steel Educational Council, 
Moraga, California. (available at www.aisc.org). 
Berman, J., Bruneau, M. (2003). Plastic Analysis and Design of Steel Plate Shear Walls, 
ASCE Journal of Structural Engineering, (in press). 
Caccese, V., Elgaaly, M., and Chen, R., (1993), Experimental Study of Thin Steel-Plate Shear 
Walls Under Cyclic Load, Journal of Structural Engineering, ASCE, Vol. 119, No. 2, Feb. 
1993, pp. 573-587. 
Clark, P., Aiken, I., Kasai, K., Ko, E., and Kimura, I., 1999, "Design Procedures for Buildings 
Incorporating Hysteretic Damping Devices," Proceedings 68th Annual Convention, pp. 355- 
371, Structural Engineers Association of California, Sacramento, CA. 
Clark, P., Kasai, K., Aiken, I., Kimura, I., 1999, Evaluation of Simplified Design Approaches 
for Structures with Yielding Steel Braces for Energy Dissipation, Proceedings of the 1999 
Annual Meeting of the Los Angeles Tall Buildings Structural Design Council, Los Angeles, 
CA. 
Driver, R.G., Kulak, G.L., Kennedy, D.J.L., and Elwi, A.E. (1997), Seismic Behavior of Steel 
Plate Shear Walls, Structural Engineering Report No. 215, Department of Civil Engineering, 
University of Alberta, Edmonton, Alberta, Canada. 
Elgaaly M. (1998), Thin Steel Plate Shear Walls Behavior and Analysis, Thin Walled 
Structures, Vol. 32, pp. 151-180. 
Elgaaly, M., Caccese, V., and Du, C. (1993), Postbuckling Behavior of Steel-Plate Shear 
Walls Under Cyclic Loads, Journal of Structural Engineering, ASCE, Vol. 119, No. 2, Feb. 
1993, pp. 588-605. 
Elgaaly, M., and Lui, Y., (1997), Analysis of Thin-Steel-Plate Shear Walls, Journal of 
Structural Engineering, ASCE, Vol. 123, No. 11, Nov. 1997, pp 1487-1496. 
Kalyanaraman, V., Sridhara, B.N., Mahadevan, K., 1994, Sleeved Column System, 
Proceedings of the SSRC Task Group Meetings and Task Force Sessions, Lehigh University, 
Bethlehem, PA. 
2003 Commentary, Chapter 8 
196 
Kalyanaraman, V., Sridhara, B.N., Thairani, V., 1998, Core Loaded Earthquake Resistant 
Bracing Design, Proceedings of the 2nd International Conference on Steel Construction, San 
Sebastian, Spain. 
Kalyanaraman, V., Sridhara, B.N., Mahadevan, K., Thairani, V., and Ramachandrran, B., 1988, 
A Sleeved Bracing System for Earthquake Resistant Design of Tall Buildings, Symposium on 
Earthquake Engineering, Univ. of Roorkee Dec. 17-19, 1998. 
Kasai, K., Fu, Y., and Watanabe, A., 1998, "Passive Control Systems for Seismic Damage 
Mitigation," Journal of Structural Engineering, Vol. 124, No. 5, (May), pp. 501-512, 
American Society of Civil Engineers, Reston, VA. 
Lubell, A.S., Prion, H.G.L., Ventura, C.E., and Rezai, M. (2000), Unstiffened Steel Plate 
Shear Wall Performance Under Cyclic Loading, Journal of Structural Engineering, ASCE, 
Vol. 126, No.4, April 2000, pp. 453-460. 
Prasad, B. 1992, Experimental Investigation of Sleeved Column, Proceedings of the 33rd 
Structural Dynamics and Materials Conference, American Institute of Aeronautics and 
Astronautics, Dallas. 
Reina, P., and Normile, D., 1997, "Fully Braced for Seismic Survival," Engineering News- 
Record, July 21, pp. 34-36, The McGraw Hill Companies, New York, NY. 
Rezai, M. (1999), Seismic Behavior of Steel Plate Shear Walls by Shake Table Testing, 
Ph.D. Dissertation, University of British Columbia, Vancouver, British Columbia, Canada. 
Roberts, T.M., and Sabouri-Ghomi, S. (1992), Hysteretic Characteristics of Unstiffened 
Perforated Steel Plate Shear Walls, Thin Walled Structures, Vol. 14, pp. 139-151. 
Sabelli, R., 2001, Research on Improving the Design and Analysis of Earthquake-Resistant 
Steel Braced Frames, EERI/FEMA NEHRP Fellowship Report, Oakland, California. 
Sabouri-Ghomi, S., and Roberts, T.M. (1992), Nonlinear Dynamic Analysis of Steel Plate 
Shear Walls Including Shear and Bending Deformations, Engineering Structures, Vol. 14, No. 
3, pp. 309-317. 
Seismic Isolation Engineering for Nippon Steel Corporation, 1999, Tests of Nippon Steel 
Corporation Unbonded Braces. Report to Ove Arup & Partners, California. 
Seismic Isolation Engineering, 1999, Report on Pre-Prototype Tests of Yielding Braces, 
California. 
Serrette, R. 1996. Shear Wall Values for Light Weight Steel Framing. American Iron and 
Steel Institute. 
Sridhara, B.N., 1990, Sleeved Column-As a Basic Compression Member, 4th International 
Conference of Steel Structures and Space Frames, Singapore. 
Thorburn, L.J., Kulak, G.L., and Montgomery, C.J. (1983), Analysis of Steel Plate Shear 
Walls, Structural Engineering Report No. 107, Department of Civil Engineering, Universtiy 
of Alberta, Edmonton, Alberta, Canada. 
Timler, P.A. and Kulak, G.L. (1983), Experimental Study of Steel Plate Shear Walls, 
Structural Engineering Report No. 114, Department of Civil Engineering, University of 
Alberta, Edmonton, Alberta, Canada. 
Tremblay, R., Degrange, G., and Blouin, J., 1999, Seismic Rehabilitation of a Four-Storey 
Building with a Stiffened Bracing System, Proc. 8th Can. Conf. on Earthquake Engrg., 
Canadian Association for Earthquake Engineering, Vancouver, B.C., 549-554. 
Steel Structure Design Requirements 
197 
Tromposch, E.W., and Kulak, G.L. (1987), Cyclic and Static Behaviour of Thin Panel Steel 
Plate Shear Walls, Structural Engineering Report No. 145, Department of Civil Engineering, 
University of Alberta, Edmonton, Alberta, Canada. 
Wada, A., Connor, J., Kawai, H., Iwata, M., and Watanabe, A., 1994, "Damage Tolerant 
Structure," ATC-15-4 Proceedings of Fifth US-Japan Workshop on the Improvement of 
Building Structural Design and Construction Practices, pp. 27-39, Applied Technology 
Council, Redwood City, CA. 
Wada, A., Saeki, E., Takeuchi, T., and Watanabe, A., c. 1998, "Development of Unbonded 
Brace," Nippon Steel's Unbonded Braces (promotional document), pp. 1-16, Nippon Steel 
Corporation Building Construction and Urban Development Division, Tokyo, Japan. 
Watanabe, A., Hitomi Y., Saeki, E., Wada, A., and Fujimoto, M., 1988, "Properties of Brace 
Encased in Buckling-Restraining Concrete and Steel Tube," Proceedings of Ninth World 
Conference on Earthquake Engineering, Vol. IV, pp. 719-724, Japan Association for 
Earthquake Disaster Prevention, Tokyo-Kyoto, Japan. 

2003 Commentary, Chapter 8 
198 
This page is intentionally left blank. 

 199 
Chapter 9 Commentary 
CONCRETE STRUCTURE DESIGN REQUIREMENTS 
9.1 GENERAL 
9.1.2 References. The main concern of Chapter 9 is the proper detailing of reinforced concrete 
construction for earthquake resistance. The bulk of the detailing requirements in this chapter are 
contained in ACI 318. The commentary for ACI 318 contains a valuable discussion of the rationale 
behind detailing requirements and is not repeated here. 
9.2 GENERAL DESIGN REQUIREMENTS 
9.2.1 Classification of shear walls. In the 2000 Provisions, shearwalls are classified by the amount and 
type of detailing required. This classification was developed to facilitate assigning shearwalls to seismic 
design categories. 
9.2.2 Modifications to ACI 318. The modifications noted for ACI 318 are: changes in load factors 
necessary to coordinate with the equivalent yield basis of this document, additional definitions and 
provisions necessary for seismic design requirements for structural systems composed of precast 
elements, and changes that incorporate certain features of the detailing requirements for reinforced 
concrete that have been adopted into the 1997 Uniform Building Code and the 2000 International 
Building Code. 
Procedures for design of a seismic-force-resisting structural system composed of precast elements 
interconnected predominately by dry joints require prior acceptance testing of modules of the generic 
structural system because with the existing state-of-knowledge, it is inappropriate to propose code 
provisions without such verification. 
The complexity of structural systems, configurations, and details possible with precast concrete elements 
requires: 
1. Selecting functional and compatible details for connections and members that are reliable and can be 
built with acceptable tolerances; 
2. Verifying experimentally the inelastic force-deformation relationships for welded, bolted, or grouted 
connections proposed for the seismic resisting elements of the building; and 
3. Analyzing the building using those connection relationships and the inelastic reversed cyclic loading 
effects imposed by the anticipated earthquake ground motions. 
Research conducted to date (Cheok and Lew, 1991; Elliott et al, 1992; Englekirk, 1987; French et al, 
1989; BSSC, 1987; Hawkins and Englekirk, 1987; Jayashanker and French, 1988; Mast, 1992; Nakaki 
and Englekirk, 1991; Neille, 1977; New Zealand Society, 1991; Pekau and Hum, 1991; Powell et al, 
1993; Priestley, 1991; Priestley and Tao, 1992; Stanton et al, 1986; Stanton et al, 1991) documents 
concepts for design using dry connections and the behavior of structural systems and subassemblages 
composed of precast elements both at and beyond peak strength levels for nonlinear reversed cyclic 
loadings. 
Use of prestressing tendons. Sec. 9.2.2.1.4 defines conditions under which prestressing tendons can be 
used, in conjunction with deformed reinforcing bars, in frames resisting earthquake forces. As 
documented in Ishizuka and Hawkins (1987), if those conditions are met no modification is necessary to 
the R and Cd factors of Table 4.3-1 when prestressing is used. Satisfactory seismic performance can be 
obtained when prestressing amounts greater than those permitted by Sec. 9.2.2.1.4 are used. However, as 
documented by Park and Thompson (1977) and Thompson and Park (1980) and as required by the 
combination of New Zealand Standards 3101:1982 and 4203:1992, ensuring satisfactory performance 
requires modification of the R and Cd factors. 
2003 Commentary, Chapter 9 
200 
9.3 SEISMIC DESIGN CATEGORY B 
Special details for ductility and toughness are not required in Seismic Design Category B. 
9.3.1 Ordinary moment frames. Since ordinary frames are permitted only in Seismic Design 
Categories A and B, they are not required to meet any particular seismic requirements. Attention should 
be paid to the often overlooked requirement for joint reinforcement in Sec.11.11.2 of ACI 318. 
9.4 SEISMIC DESIGN CATEGORY C 
A frame used as part of the seismic-force-resisting system in Seismic Design Category C is required to 
have certain details that are intended to help sustain integrity of the frame when subjected to deformation 
reversals into the nonlinear range of response. Such frames must have attributes of intermediate moment 
frames. Structural (shear) walls of buildings in Seismic Design Category C are to be designed in 
accordance with the requirements of ACI 318. 
9.4.1.1 Moment frames. The concept of moment frames for various levels of hazard zones and of 
performance is changed somewhat from the provisions of ACI 318. Two sets of moment frame detailing 
requirements are defined in ACI 318, one for regions of high seismic risk and the other for regions of 
moderate seismic risk. For the purposes of this document, the regions are made equivalent to Seismic 
Design Categories in which high risk means Seismic Design Categories D and E and moderate risk 
means Seismic Design Category C. This document labels these two frames the special moment frame 
and the intermediate moment frame, respectively. 
The level of inelastic energy absorption of the two frames is not the same. The Provisions introduce the 
concept that the R factors for these two frames should not be the same. The preliminary version of the 
Provisions (ATC 3-06) assigned the R for ordinary frames to what is now called the intermediate frame. 
In spite of the fact that the R factor for the intermediate frame is less than the R factor for the special 
frame, use of the intermediate frame is not permitted in the higher Seismic Design Categories (D, E, and 
F). On the other hand, this arrangement of the Provisions encourages consideration of the more stringent 
detailing practices for the special frame in Seismic Design Category C because the reward for use of the 
higher R factor can be weighed against the higher cost of the detailing requirements. The Provisions also 
introduce the concept that an intermediate frame may be part of a dual system in Seismic Design 
Category C. 
The differences in the performance basis of the requirements for the two types of frames might be 
summarized briefly as follows (see the commentary of ACI 318 for a more detailed discussion of the 
requirement for the special frame): 
1. The shear strength of beams and columns must not be less than that required when the member has 
yielded at each end in flexure. For the special frame, strain hardening and other factors are 
considered by raising the effective tensile strength of the bars to 125 percent of specified yield. For 
the intermediate frame, an escape clause is provided in that the calculated shear using double the 
prescribed seismic force may be substituted. Both types require the same minimum amount and 
maximum spacing of transverse reinforcement throughout the member. 
2. The shear strength of joints is limited and special provisions for anchoring bars in joints exist for 
special moment frames but not intermediate frames. Both frames require transverse reinforcement in 
joints although less is required for the intermediate frame. 
3. Closely spaced transverse reinforcement is required in regions of potential hinging (typically the ends 
of beams and columns) to control lateral buckling of longitudinal bars after the cover has spalled. 
The spacing limit is slightly more stringent for columns in the special frame. 
4. The amount of transverse reinforcement in regions of hinging for special frames is empirically tied to 
the concept of providing enough confinement of the concrete core to preserve a ductile response. 
Concrete Structure Design Requirements 
201 
These amounts are not required in the intermediate frame and, in fact, for beams stirrups may be used 
in lieu of hoops. 
5. The special frame must follow the strong column/weak beam rule. Although this is not required for 
the intermediate frame, it is highly recommended for multistory construction. 
6. The maximum and minimum amounts of reinforcement are limited to prevent rebar congestion and to 
assure a nonbrittle flexural response. Although the precise limits are different for the two types of 
frames, a great portion of practical, buildable designs will satisfy both. 
7. Minimum amounts of continuous reinforcement to account for moment reversals are required by 
placing lower limits on the flexural strength at any cross section. Requirements for the two types of 
frames are similar. 
8. Locations for splices of reinforcement are more tightly controlled for the special frame. 
9. In addition, the special frame must satisfy numerous other requirements beyond the intermediate 
frame to assure that member proportions are within the scope of the present research experience on 
seismic resistance and that analysis, design procedures, qualities of the materials, and inspection 
procedures are at the highest level of the state of the art. 
9.5 SEISMIC DESIGN CATEGORIES D, E, AND F 
The requirements conform to current practice in the areas of highest seismic hazard. 
9.6 ACCEPTANCE CRITERIA FOR SPECIAL PRECAST STRUCTURAL WALLS 
BASED ON VALIDATION TESTING 
9.6.1. Notation 
Symbols additional to those in Chapter 21 of ACI 318 are defined: 
Ah = area of hysteresis loop 
E1,E2 = peak lateral resistance for positive and negative loading, respectively, for third cycle of loading 
sequence. 
f1 = live load factor defined in 9.6.2.3. 
hw = height of column of test module, in. or mm. 
K, K = initial stiffness for positive and negative loading, respectively, for first cycle 
.1,.2 = drift ratios at peak lateral resistance for positive and negative loading, respectively, for third 
cycle of loading sequence.. 
.1',.2' = drift ratios for zero lateral load for unloading at stiffness K, K from peak positive and 
negative lateral resistance, respectively, for third cycle of loading sequence. (Figure C9.6.2.4) 
. = lateral displacement, in. or mm. See Figures. C9.6.2.2.1, C9.6.2.2.2 and C9.6.2.2.3 
.a = allowable story drift, in. or mm. See Table 9.5.2.8 of SEI/ASCE 7-02 
9.6.2 Definitions 
9.6.2.1 Coupling elements. Coupling elements are connections provided at specific intervals along the 
vertical boundaries of adjacent structural walls. Coupled structural walls are stiffer and stronger than the 
same walls acting independently. For cast-in-place construction effective coupling elements are typically 
coupling beams having small span-to-depth ratios. The inelastic behavior of such beams is normally 
controlled by their shear strength. For precast construction, effective coupling elements can be precast 
beams connected to the adjacent structural walls either by post-tensioning, ductile mechanical devices, or 
grouted-in-place reinforcing bars. The resultant coupled construction can be either emulative of cast-in-
2003 Commentary, Chapter 9 
202 
place construction or non-emulative(jointed). However, for precast construction coupling beams can also 
be omitted and mechanical devices used to connect directly the vertical boundaries of adjacent structural 
walls2,3. 
9.6.2.2 Drift ratio. The definition of the drift ratio, ., is illustrated in Figure C9.6.2.2.1 for a three panel 
wall module. The position of the module at the start of testing, with only its self-weight acting, is 
indicated by broken lines. The module is set on a horizontal foundation support that is centered at A and 
is acted on by a lateral force H applied at the top of the wall. The self-weight of the wall is distributed 
uniformly to the foundation support. However, under lateral loading, that self-weight and any axial 
gravity load acting at the top of the wall cause overturning moments on the wall that are additional to the 
overturning moment Hhw and can affect deformations. The chord AB of the centroidal axis of the wall is 
the vertical reference line for drift measurements. 
For acceptance testing a lateral force H is applied to the wall through the pin at B. Depending on the 
geometric and reinforcement characteristics of the module that force can result in the module taking up 
any one, or a combination, of the deformed shapes indicated by solid lines in Figures C9.6.2.2.1, 
C9.6.2.2.2 and C9.6.2.2.3. 

Concrete Structure Design Requirements 
203 
Figure C9.6.2.2.2 illustrates several possible components of the displacement . for a wall that is 
effectively solid while Figure C9.6.2.2.3 illustrates two possibly undesirable components of the 
displacement .. Regardless of the mode of deformation of the wall, the lateral force causes the wall at B 
to displace horizontally by an amount .. The drift ratio is the angular rotation of the wall chord with 
respect to the vertical and for the setup shown equals . / hw where hw is the wall height and is equal to the 
distance between the foundation support at A and the load point at B. 
Where prestressing steel is used in wall members, the stress fps in the reinforcement at the nominal and the 
probable lateral resistance shall be calculated in accordance with Sec. 18.7 of ACI 318. 
9.6.2.3 Global toughness. These provisions describe acceptance criteria for special precast structural 
walls based on validation testing. The requirements of Sec. 21.2.1.5 of ACI 318 concerning toughness 
cover both to the energy dissipation of the wall system which, for monolithic construction, is affected 
primarily by local plastic hinging behavior and the toughness of the prototype structure as a whole. The 
2003 Commentary, Chapter 9 
204 
latter is termed global toughness in these provisions and is a condition that does not apply to the walls 
alone. That global toughness requirement can be satisfied only though analysis of the performance of the 
prototype structure as a whole when the walls perform to the criteria specified in these provisions. 
The required gravity load for global toughness evaluations is the value given by these provisions. For 
conformity with Sec. 9.2.1 of ACI 318, UBC 1997, IBC 2003 and NFPA 5000, the required gravity load 
is 1.2D + f1L where the seismic force is additive to gravity forces and 0.9D where the seismic force 
counteracts gravity forces. D is the effect of dead loads, L is the effect of live loads, and f1 is a factor 
equal to 0.5 except for garages, areas occupied as places of public assembly, and all areas where the live 
load is greater than 100 psf (4.79 kN/m2) where f1 equals 1.0. 
9.6.2.5 Relative energy dissipation ratio. This concept is illustrated in Figure C9.6.2.5 for the third 
loading cycle to the limiting drift ratio required by Sec. 9.6.7.4, 9.6.7.5 or 9.6.7.6, as appropriate. 
For Figure C9.6.2.5, it is assumed that the test module has exhibited different initial stiffnesses, K and 
K, for positive and negative lateral forces and that the peak lateral resistances for the third cycle for the 
positive and negative loading directions, E1 and E2, also differ. The area of the hysteresis loop for the 
third cycle, Ah, is hatched. The circumscribing figure consists of two parallelograms, ABCD and DFGA. 
The slopes of the lines AB and DC are the same as the initial stiffness, K, for positive loading and the 
slopes of the lines DF and GA are the same as the initial stiffness, K', for negative loading. The relative 
energy dissipation ratio concept is similar to the equivalent viscous damping concept used in Sec. 13.9.3 
of the 2000 NEHRP Provisions and Commentary for required tests of seismic isolation systems. 
For a given cycle the relative energy dissipation ratio, , is the area, Ah, inside the lateral force-drift ratio 
loop for the module, divided by the area of the effective circumscribing parallelograms ABCD and 
DFGA. The areas of the parallelograms equal the sum of the absolute values of the lateral force strengths, 
E1 and E2, at the drift ratios .1 and .2 multiplied by the sum of the absolute values for the drift ratios .1' 
and .2'. 
9.6.3 Scope and general requirements. While only ACI Committee 318 can determine the 
requirements necessary for precast walls to meet the provisions of Sec. 21.2.1.5 of ACI 318, Sec. 1.4 of 
Concrete Structure Design Requirements 
205 
ACI 318 already permits the building official to accept wall systems, other than those explicitly covered 
by Chapter 21 of ACI 318, provided specific tests, load factors, deflection limits, construction procedures 
and other pertinent requirements have been established for acceptance of such systems consistent with the 
intent of the code. The purpose of these provisions is to provide a framework that establishes the specific 
tests, load factors, deflection limits and other pertinent requirements appropriate for acceptance, for 
regions of high seismic risk or for structures assigned to high seismic performance or design categories, 
of precast wall systems, including coupled wall systems, not satisfying all the requirements of Chapter 21 
of ACI 318. For regions of moderate seismic risk or for structures assigned to intermediate seismic 
performance or design categories, less stringent provisions than those specified here are appropriate. 
These provisions assume that the precast wall system to be tested has details differing from those 
prescribed by Sec. 21.7 of ACI 318 for conventional monolithic reinforced concrete construction. Such 
walls may, for example, involve the use of precast elements, precast prestressed elements, post-tensioned 
reinforcement, or combinations of those elements and reinforcement. 
For monolithic reinforced concrete walls a fundamental design requirement of Chapter 21 of ACI 318 is 
that walls with hw/lw exceeding1.0 be proportioned so that their inelastic response is dominated by flexural 
action on a critical section located near the base of the wall. That fundamental requirement is retained in 
these provisions. The reason is that tests on modules, as envisioned in these provisions, cannot be 
extrapolated with confidence to the performance of panelized walls of proportions differing from those 
tested for the development of Chapter 21of ACI 318 if the shear-slip displacement pattern of Figure 
C9.6.2.2.3, or the shear deformation response of Figure C9.6.2.2.2, governs the response developed in the 
test on the module. Two other fundamental requirements of Chapter 21 of ACI 318 are for ties around 
heavily strained boundary element reinforcement and the provision of minimum amounts of uniformly 
distributed horizontal and vertical reinforcement in the web of the wall. Ties around boundary element 
reinforcement to inhibit its buckling in compression are required where the strain in the extreme 
compression fiber is expected to exceed some critical value. Minimum amounts of uniformly distributed 
horizontal and vertical reinforcement over the height and length of the wall are required to restrain the 
opening of inclined cracks and allow the development of the drift ratios specified in Sec. 9.6.7.4, 9.6.7.5 
and 9.6.7.6. Deviations from those tie and distributed reinforcement requirements are possible only if a 
theory is developed that can substantiate reasons for such deviations and that theory is tested as part of 
the validation testing. 
9.6.3.1. These provisions are not intended for use with existing construction or for use with walls that are 
designed to conform to all the requirements of Sec. 21.7 of ACI 318. The criteria of these provisions are 
more stringent than those for walls designed to Sec. 21.7 of ACI 318. Some walls designed to 21.7, and 
having low height to length ratios, may not meet the drift ratio limits of Eq. 9.6.1 because their behavior 
may be governed by shear deformations. The height to length ratio of 0.5 is the least value for which Eq. 
9.6.1 is applicable. 
9.6.3.3. For acceptance, the results of the tests on each module must satisfy the acceptance criteria of 
Sec. 9.6.9. In particular, the relative energy dissipation ratio calculated from the measured results for the 
third cycle between the specified limiting drift ratios must equal or exceed 1/8. For uncoupled walls, 
relative energy dissipation ratios increase as the drift ratio increases. Tests on slender monolithic walls 
have shown relative energy dissipation ratios, derived from rotations at the base of the wall, of about 40- 
45 percent at large drifts. The same result has been reported even where there has been a significant 
opening in the web of the wall on the compression side. For 0.020 drift ratios and walls with height to 
length ratios of 4, relative energy dissipation ratios have been computed as 30, 18, 12, and 6 percent, for 
monolithic reinforced concrete, hybrid reinforced/post-tensioned prestressed concrete with equal flexural 
strengths provided by the prestressed and deformed bar reinforcement, hybrid reinforced/post-tensioned 
prestressed concrete with 25 percent of the flexural strength provided by deformed bar reinforcement and 
75 percent by the prestressed reinforcement, and post-tensioned prestressed concrete special structural 
walls, respectively. Thus, for slender precast uncoupled walls of emulative or non-emulative design it is 
to be anticipated that at least 35 percent of the flexural capacity at the base of the wall needs to be 
2003 Commentary, Chapter 9 
206 
provided by deformed bar reinforcement if the requirement of a relative energy dissipation ratio of 1/8 is 
to be achieved. However, if more than about 40 percent of the flexural capacity at the base of the wall is 
provided by deformed bar reinforcement, then the self-centering capability of the wall following a major 
event is lost and that is one of the prime advantages gained with the use of post-tensioning. For squat 
walls with height to length ratios between 0.35 and 0.69 the relative energy dissipation has been 
reported13 as remaining constant at 23 percent for drifts between that for first diagonal cracking and that 
for a post-peak capacity of 80 percent of the peak capacity. Thus, regardless of whether the behavior of a 
wall is controlled by shear or flexural deformations a minimum relative energy dissipation ratio of 1/8 is a 
realistic requirement. 
For coupled wall systems, theoretical studies and tests have demonstrated that the 1/8 relative energy 
dissipation ratio can be achieved by using central post-tensioning only in the walls and appropriate 
energy dissipating coupling devices connecting adjacent vertical wall boundaries. 
9.6.3.3.4. The SEI/ASCE 7-02 allowable story drift limits are the basis for the drift limits of IBC 2003 
and NFPA 5000. Allowable story drifts, .a, are specified in Table 1617.3 of IBC 2003 and likely values 
are discussed in the Commentary to Sec. 9.6.7.4. The limiting initial drift ratio consistent with .a equals 
.a/fCdhw, where f is the strength reduction factor appropriate to the condition, flexure or shear, that 
controls the design of the test module. For example, for .a/hw equal to 0.015, the required deflection 
amplification factor Cd of 5, and f equal to 0.9, the limiting initial drift ratio, corresponding to B in 
Figure C9.6.9.1, is 0.0033. The use of a f value is necessary because the allowable story drifts of the IBC 
are for the design seismic load effect, E, while the limiting initial drift ratio is at the nominal strength, En , 
which must be greater than E/f. The load-deformation relationship of a wall becomes significantly nonlinear 
before the applied load reaches Ent. While the load at which that non-linearity becomes marked 
depends on the structural characteristics of the wall, the response of most walls remains linear up to about 
75 percent of Ent. 
9.6.3.3.5. The criteria of Sec. 9.6.9 are for the test module. In contrast, the criterion of Sec. 9.6.3.3.5 is 
for the structural system as a whole and can be satisfied only by the philosophy used for the design and 
analysis of the building as a whole. The criterion adopted here is similar to that described in the last 
paragraph of R21.2.1 of ACI 318 and the intent is that test results and analyses demonstrate that the 
structure, after cycling three times through both positive and negative values of the limiting drift ratio 
specified in Sec. 9.6.7.4, 9.6.7.5 or 9.6.7.6, as appropriate, is still capable of supporting the gravity load 
specified as acting on it during the earthquake. 
9.6.4 Design procedure 
9.6.4.1. The test program specified in these provisions is intended to verify an existing design procedure 
for precast structural walls for a specific structure or for prequalifying a generic type of special precast 
wall system for construction in general. The test program is not for the purpose of creating basic 
information on the strength and deformation properties of such systems for design purposes. Thus, the 
test modules should not fail during the validation testing, a result that is the opposite of what is usually 
necessary during testing in the development phase for a new or revised design procedure. For a generic 
precast wall system to be accepted based on these provisions, a rational design procedure is to have been 
developed prior to this validation testing. The design procedure is to be based on a rational consideration 
of material properties and force transfer mechanisms, and its development will usually require 
preliminary and possibly extensive physical testing that is not part of the validation testing. Because 
special wall systems are likely to respond inelastically during design-level ground shaking, the design 
procedure must consider wall configuration, equilibrium of forces, compatibility of deformations, the 
magnitudes of the lateral drifts, reversed cyclic displacements, the relative values of each limiting 
engineering design criteria (shear, flexure and axial load) and use appropriate constitutive laws for 
materials that include considerations of effects of cracking, loading reversals and inelasticity. 
The effective initial stiffness of the structural walls is important for calculating the fundamental period of 
the prototype structure. The procedure used to determine the effective initial stiffness of the walls is to be 
Concrete Structure Design Requirements 
207 
verified from the validation test results as described in Sec. 9.6.7.11. 
Provisions Sec. 9.6.4.1.1 through 9.6.4.1.3 state the minimum procedures to be specified in the design 
procedure prior to the start of testing. The Authority Having Jurisdiction may require that more details be 
provided in the design procedure than those of Sec. 9.6.4.1.1 through 9.6.4.1.3 prior to the start of testing. 
9.6.4.2. The justification for the small number of test modules, specified in Sec. 9.6.5.1 is that a previously 
developed rational design procedure is being validated by the test results. Thus, the test modules for the 
experimental program must be designed using the procedure intended for the prototype wall system and strengths 
must be predicted for the test modules before the validation testing is started. 
9.6.5 Test modules. 
9.6.5.1. One module must be tested for each limiting engineering design criterion, such as shear, or axial 
load and flexure, for each characteristic configuration of walls. Thus, in accordance with 9.6.4.3 if the 
test on the module results in a maximum shear stress of 3 c f ' then the maximum shear stress that can 
be used in the prototype is that same value. Each characteristic in-plane configuration of walls, or coupled 
walls, in the prototype structure must also be tested. Thus, as a minimum for one-way structural walls, 
two modules with the configuration shown in Figure C9.6.2.2.1, and, for one way coupled walls, two 
modules with the configuration shown in either Figure C9.6.5.1(a) or in Figure C9.6.5.1(b), must be 
tested. In addition, if intersecting wall systems are to be used then the response of the wall systems for 
the two orthogonal directions needs to be tested. For two-way wall systems and coupled wall-frame 
systems, testing of configurations other than those shown in Figures C9.6.2.2.1 and C9.6.5.1 may be 
appropriate when it is difficult to realistically model the likely dominant earthquake deformations using 
orthogonal direction testing only. 
This provision should not be interpreted as implying that only two tests will need to be made to qualify a 
generic system. During the development of that system it is likely that several more tests will have been 
made, resulting in progressive refinements of the mathematical model used to describe the likely 
performance of the generic structural wall system and its construction details. Consequently, only one 
test of each module type for each limiting engineering design condition, at a specified minimum scale and 
subjected to specific loading actions, may be required to validate the system. Further, as stated in Sec. 
9.6.9.1, if any one of those modules for the generic wall system fails to pass the validation testing 
2003 Commentary, Chapter 9 
208 
required by these provisions, then the generic wall system has failed the validation testing 
In most prototype structures, a slab is usually attached to the wall and, as demonstrated by the results of 
the PRESSS building test, the manner in which the slab is connected to the wall needs to be carefully 
considered. The connection needs to be adequate to allow the development of story drifts equal to those 
anticipated in these provisions. However, in conformity with common practice for the sub-assemblage 
tests used to develop the provisions of Chapter 21of ACI 318, there is no requirement for a slab to be 
attached to the wall of the test module. The effect of the presence of the slab should be examined in the 
development program that precedes the validation testing. 
9.6.5.3. Test modules need not be as large as the corresponding walls in the prototype structure. The 
scale of the test modules, however, must be large enough to capture all the complexities associated with 
the materials of the prototype wall, its geometry and reinforcing details, load transfer mechanisms, and 
joint locations. For modules involving the use of precast elements, for example, scale effects for load 
transfer through mechanical connections should be of particular concern. The issue of the scale 
necessary to capture fully the effects of details on the behavior of the prototype should be examined in the 
development program that precedes the validation testing. 
9.6.5.4. It is to be expected that for a given generic precast wall structure, such as an unbonded centrally 
post-tensioned wall constructed using multiple precast or precast pretensioned concrete wall panels, 
validation testing programs will initially use specific values for the specified strength of the concrete and 
reinforcement in the walls, the layout of the connections between panels, the location of the posttensioning, 
the location of the panel joints, and the design stresses in the wall. Pending the development 
of an industry standard for the design of such walls, similar to the standard for special hybrid moment 
frames, specified concrete strengths, connection layouts, post-tensioning amounts and locations, etc., used 
for such walls will need to be limited to the values and layouts used in the validation testing programs. 
9.6.5.5. For walls constructed using precast or precast/prestressed panels and designed using nonemulative 
methods, the response under lateral load can change significantly with joint opening (Figure 
C9.6.2.2.2(d) and Figure C9.6.2.2.3(a)). The number of panels used to construct a wall depends on wall 
height and design philosophy. If, in the prototype structure, there is a possibility of horizontal joint 
opening under lateral loading at a location other than the base of the wall, then the consequences of that 
possibility need to be considered in the development and validation test programs. Joint opening at 
locations other than the base can be prevented through the use of capacity design procedures. 
9.6.5.6. The significance of the magnitude of the gravity load that acts simultaneously with the lateral 
load needs to be addressed during the validation testing if the development program suggests that effect is 
significant. 
9.6.5.7. Details of the connection of walls to the foundation are critical, particularly for non-emulative 
wall designs. The deformations that occur at the base of the wall due to plastic hinging or extension of 
the reinforcing bars or post-tensioning steel crossing the wall to foundation interface, (Figure 
C9.6.2.2.2(d)), are in part determined by details of the anchorage and the bonding of those reinforcements 
on either side of the interface. Grout will be normally used to bed panels on the foundation and the 
characteristics of that grout in terms of materials, strength and thickness, can have a large effect on wall 
performance. The typical grout pad with a thickness of 1 inch (25 mm) or less can be expected to provide 
a coefficient of friction of about 0.6 under reversed loadings. Pads with greater thickness and without 
fiber reinforcement exhibit lesser coefficients of friction. Adequate frictional resistance is essential to 
preventing undesirable shear-slip deformations of the type shown in Figure C9.6.2.2.3(b). 
9.6.5.8. The geometry of the foundations need not duplicate that used in the prototype structure. 
However, the geometric characteristics of the foundations (width, depth and length) need to be large 
enough that they do not influence the behavior of the test module. 
9.6.6 Testing agency. In accordance with the spirit of the requirements of Sec.1.3.5 and 1.4 of ACI 318, 
it is important that testing be carried out by a recognized independent testing agency, approved by the 
Concrete Structure Design Requirements 
209 
agency having jurisdiction and that the testing and reporting be supervised by a registered design 
professional familiar with the proposed design procedure and experienced in testing and seismic 
structural design. 
9.6.7 Test method. The test sequence is expressed in terms of drift ratio, and the initial ratio is related to 
the likely range of linear elastic response for the module. That approach, rather than testing at specific 
drift ratios of 0.005, 0.010, etc., is specified because, for modules involving prestressed concrete, the 
likely range of elastic behavior varies with the prestress level. 
An example of the test sequence specified in Sec. 9.6.7.2 through 9.6.7.6 is illustrated in Figure C9.6.7. 
The sequence is intended to ensure that displacements are increased gradually in steps that are neither too 
large nor too small. If steps are too large, the drift capacity of the system may not be determined with 
sufficient accuracy. If the steps are too small, the system may be unrealistically softened by loading 
repetitions, resulting in artificially low maximum lateral resistances and artificially high maximum drifts. 
Also, when steps are too small, the rate of change of energy stored in the system may be too small 
compared with the change occurring during a major event. Results, using such small steps, can mask 
undesirable brittle failure modes that might occur in the inelastic response range during a major event. 
Because significant diagonal cracking is to be expected in the inelastic range in the web of walls, and in 
particular in squat walls, the pattern of increasing drifts used in the test sequence can markedly affect 
diagonal crack response in the post-peak range of behavior. 
The drift capacity of a building in a major event is not a single quantity, but depends on how that event 
shakes the structure. In the forward near field, a single pulse may determine the maximum drift demand, 
in which case a single large drift demand cycle for the test module would give the best estimation of the 
drift capacity. More often, however, many small cycles precede the main shock and that is the scenario 
represented by the specified loading. 
There is no requirement for an axial load to be applied to the wall simultaneously with the application of 
the lateral displacements. In many cases it will be conservative not to apply axial load because, in 
general, the shear capacity of the wall and the resistance to slip at the base of the wall increase as the axial 
load on the wall increases. However, as the height of the wall increases and the limiting drift utilized in 
the design of the wall increases, the likelihood of extreme fiber crushing in compression at maximum 
drift increases, and the importance of the level of axial load increases. The significance of the level of 
2003 Commentary, Chapter 9 
210 
axial loading should be examined during the development phase. 
9.6.7.4. For the response of a structure to the design seismic shear force, current building codes such as 
UBC-97, IBC 2003 or NFPA 5000 , or recommended provisions such as 2000 Provisions, SEI/ASCE 7- 
02 and FEMA 273 specify a maximum allowable drift. However, structures designed to meet that drift 
limit may experience greater drifts during an earthquake equal to the design basis earthquake and are 
likely to experience greater drifts during an earthquake equal to the maximum credible earthquake. In 
addition to the characteristics of the ground motion, actual drifts will depend on the strength of the 
structure, its initial elastic stiffness, and the ductility expected for the given lateral load resisting system. 
Specification of suitable limiting drifts for the test modules requires interpretation and allowance for 
uncertainties in the assumed ground motions and structural properties. 
In IBC 2003, the design seismic shear force applied at the base of a building is related directly to its 
weight and the design elastic response acceleration, and inversely to a response modification factor, R. 
That R factor increases with the expected ductility of the lateral force resisting system of the building. 
Special structural walls satisfying the requirements of Sec. 21.2 and 21.7 are assigned an R value of 6 
when used in a building frame system and a value of 5 when used in a bearing wall system. They are also 
assigned allowable story drift ratios that are dependent on the hazard to which the building is exposed. 
When the design seismic shear force is applied to a building, the building responds inelastically and the 
resultant computed drifts, (the design story drifts), must be less than a specified allowable drift. 
Additional guidance is given in FEMA 356 where the deformations for rectangular walls with height to 
length ratios greater than 2.5, and flanged wall sections with height to length ratios greater than 3.5, are to 
be assumed to be controlled by flexural actions. When structural walls are part of a building representing 
a substantial hazard to human life in the event of a failure, the allowable story drift ratio for shear 
controlled walls is 0.0075 and for flexure controlled walls is a function of the plastic hinge rotation at the 
base of the wall. For flexure controlled walls values range up to a maximum of about 0.02 for walls with 
confined boundary elements with low reinforcement ratios and shear stress less than 3 c f ' 
To compensate for the use of the R value, IBC Sec. 1617.4.6 requires that the drift determined by an 
elastic analysis for the code-prescribed seismic forces be multiplied by a deflection amplification factor, 
Cd ,to determine the design story drift and that the design story drift must be less than the allowable story 
drift. In building frame systems, structural walls satisfying the requirements of Sec. 21.7of ACI 318 are 
assigned a Cd value of 5. However, research8 has found that design story drift ratios determined in the 
foregoing manner may be too low. Drift ratios of 6 times IBC-calculated values, (rather than 5), are more 
representative of the upper bounds to expected drift ratios. The value of 6 is also in agreement with the 
finding that the drift ratio of an inelastic structure is approximately the same as that of an elastic structure 
with the same initial period. For flexure controlled walls the value of 6/5 times the present IBC limits on 
calculated drift ratio, would lead to a limit on real drift ratios of up to 0.024. 
Duffy et al. reviewed experimental data for shear walls to define post-peak behavior and limiting drift 
ratios for walls with height to length ratios between 0.25 and 3.5. Seo et al. re-analyzed the data of Duffy 
et al. together with data from tests conducted subsequent to the analysis of Duffy et al. Duffy et al. 
established that for squat walls with web reinforcement satisfying ACI 318-02 requirements and height to 
length ratios between 0.25 and 1.1, there was a significant range of behavior for which drifts were still 
reliable in the post-peak response region. Typically the post-peak drift increased by 0.005 for a 20 
percent degradation in capacity under cyclic loading. For greater values of degradation, drifts were less 
reliable. That finding has also been confirmed through tests conducted by Hidalgo et al.13 on squat walls 
with effective height to length ratios ranging between 0.35 and 1.0. Values of the drift ratio of the walls 
at inclined cracking and at peak capacity varied little with web reinforcement. By contrast, drifts in the 
post-peak range were reliable to a capacity equal to 80 percent of the peak capacity and were 0.005 
greater than the drifts at peak capacity provided the walls contained horizontal and vertical web 
reinforcement equal to 0.25 percent. 
From an analysis of the available test data, and from theoretical considerations for a wall rotating 
Concrete Structure Design Requirements 
211 
flexurally about a plastic hinge at its base, Seo et al concluded that the limiting drift at peak capacity 
increased almost linearly with the height to length ratio of the wall. When the additional post peak drift 
capacity for walls with adequate web reinforcement was added to the drift at peak capacity, then the total 
available drift capacity in percent was given by the following equation: 
1.0=0.67[hw lw]+0.5=3.0 
where hw is the height of the wall, and lw is the length of the wall. The data from the tests of Hidalgo et 
al. suggest that while that formula is correct for squat walls the lower limit on drift can be decreased to 
0.8 as specified in these provisions and that the use of that formula should be limited to walls with height 
to length ratios equal to or greater than 0.5. For wall height to length ratios less than 0.5 the behavior is 
controlled principally by shear deformations, (Figure C9.6.2.2.2(c)), and Eq. 9.6.1 should not be used. 
The upper value of 0.030 for the drift ratio was somewhat optimistic because the data were for walls with 
height to length ratios equal to or less than 3.5 and subsequent tests have shown that the upper limit of 
2.5, as specified in Eq. 9.6.1, is a more realistic limit. 
9.6.7.5. The design capacity for coupled wall systems must be developed by the drift ratio corresponding 
to that for the wall with the least hw/lw value. However, it is desirable that testing be continued to the drift 
given by Eq. 9.6.1 for the wall with the greatest hw/lw in order to assess the reserve capacity of the coupled 
wall system. 
9.6.7.6. The drift limits of Eq. 9.6.1 are representative of the maximum that can be achieved by walls 
designed to ACI 318. The use of smaller drift limits is appropriate if the designer wishes to use 
performance measures less than the maximum permitted by ACI 318. Examples are the use of reduced 
shear stresses so that the likelihood of diagonal cracking of the wall is minimized or reduced compressive 
stresses in the boundary elements of the wall so that the risk of crushing is reduced. Non-linear time 
history analyses for the response to a suite of maximum considered earthquake (MCE) ground motions, 
rather than 1.5 times a suite of the corresponding design basis earthquake (DBE) ground motions, is 
required because the drifts for the response to the MCE motion can be significantly larger than 1.5 times 
the drifts for the response to the DBE motions. 
9.6.7.10. In many cases, data additional to the minimum specified in Sec. 9.6.7.7 may be useful to 
confirm both design assumptions and satisfactory response. Such data include relative displacements, 
rotations, curvatures, and strains. 
9.6.8 Test report. 
The test report must be sufficiently complete and self-contained for a qualified expert to be satisfied that 
the tests have been designed and carried out in accordance with these criteria, and that the results satisfy 
the intent of these provisions. Sec.9.6.8.1.1 through 9.6.8.1.11 state the minimum evidence to be 
contained within the test report. The Authority having Jurisdiction or the registered design professional 
supervising the testing may require that additional test information be reported. 
9.6.9 Test module acceptance criteria. 
The requirements of this clause apply to each module of the test program and not to an average of the 
results of the program. Figure C9.6.9.1 illustrates the intent of this clause. 
2003 Commentary, Chapter 9 
212 
9.6.9.1.1. Where nominal strengths for opposite loading directions differ, as is likely for C-, L- or Tshaped 
walls, the criterion of Sec. 9.6.9.1.1 applies separately to each direction. 
9.6.9.1.2. At high cyclic-drift ratios, strength degradation is inevitable. To limit the level of degradation 
so that drift ratio demands do not exceed anticipated levels, a maximum strength degradation of 0.20Emax 
is specified. Where strengths differ for opposite loading directions, this requirement applies 
independently to each direction. 
9.6.9.1.3. If the relative energy dissipation ratio is less than 1/8, there may be inadequate damping for the 
building as a whole. Oscillations may continue for some time after an earthquake, producing low-cycle 
fatigue effects, and displacements may become excessive. 
If the stiffness becomes too small around zero drift ratio, the structure will be prone to large 
displacements for small lateral force changes following a major earthquake. A hysteresis loop for the 
third cycle between peak drift ratios of 1/10 times the limiting drift ratio given by Eq. 9.6.1, that has the 
form shown in Figure C9.6.9.1, is acceptable. At zero drift ratio, the stiffnesses for positive and negative 
loading are about 11 percent of the initial stiffnesses. Those values satisfy Sec. 9.6.9.1.3. An 
unacceptable hysteresis loop form would be that shown in Figure C9.6.9.1.3 where the stiffness around 
zero drift ratio is unacceptably small for both positive and negative loading. 

Concrete Structure Design Requirements 
213 
Figure C 9.6.9.1.3 Unacceptable hysteretic behavior 
REFERENCES 
ACI Innovation Task Group I and Collaborators. 2001. Acceptance Criteria for Moment Frames Based on 
Structural Testing (T1.1-01) and Commentary(T1.1R-01). American Concrete Institute, Farmington Hills, 
MI, 2001, 10p. 
ACI Innovation Task Group I and Collaborators. 2001  Special Hybrid Moment Frames Composed of 
Discretely Jointed Precast and Post-Tensioned Concrete Members (ACI T1.2-XX) and Commentary (ACI 
T1.2R-XX), ACI Structural Journal,.98 (5):771-784. 
Ali, A. and J.K. Wight. 1990. Reinforced Concrete Structural Walls with Staggered Opening 
Configurations Under Reversed Cyclic Loading, Report UMCE 90-05, Department of Civil Engineering, 
University of Michigan, Ann Arbor, MI. 
Building Seismic Safety Council. 1987. Guide to Use of NEHRP Recommended Provisions in 
Earthquake Resistant Design of Buildings, 1985 Edition. Washington, D.C.: FEMA. 
Building Seismic Safety Council, NEHRP Recommended Provisions for Seismic Regulations for New 
Buildings and Other Structures, 2000. FEMA 368 and 369. Washington, D.C. 
Cheok, G. S., and H. S. Lew. 1991. Performance of Precast Concrete Beam-to-Column Connections 
Subject to Cyclic Loading, PCI Journal, 36, (3):56-67. 
Duffy, T.A., A.Goldman, and C.R. Farrar. 1993. Shear Wall Ultimate Drift Limits, Report 
NUREG/CR-6104, LA-12649-MS. U.S. Nuclear Regularity Commission. 
Elliott, K. S., G. Davies, and W. Omar. 1992. Experimental and Theoretical Investigation of Precast 
Concrete Hollow-Cored Slabs Used as Horizontal Floor Diaphragms, The Structural Engineer, 70, 
(10):175-187. 
Englekirk, R. E. 1987. Concepts for the Development of Earthquake Resistant Ductile Frames of 
Precast Concrete, PCI Journal, 32, (1). 
Federal Emergency Management Agency. 2000. NEHRP Guidelines for the Seismic Rehabilitation of 
2 003 Commentary, Chapter 9 
214 
Buildings, Chapter 6, Concrete, FEMA 356 and 357. 
French, C. W., M. Hafner, and V. Jayashanker. 1989. Connections Between Precast Elements - Failure 
Within Connection Region, ASCE Journal of Structural Engineering, 115, (12):3171-3192. 
Hawkins, N. M., and R. E. Englekirk. 1987. U.S.-Japan Seminar on P/C Concrete Construction in 
Seismic Zones, PCI Journal, 32,(2). 
Hidalgo, P.A.,C.A. Ledezma, and R.A. Jordan. 2002. Seismic Behavior of Squat Reinforced Concrete 
Shear Walls, Earthquake Spectra, 18 (2): 287-308. 
Hutchinson, R.L., S.H. Rizkalla, M. Lau, and M. Heuvel, 1991. Horizontal Post-Tensioned Connections 
for Precast Concrete Bearing Shear Walls, PCI Journal, 36,(3):.64-76. 
International Code Council. 2003. International Building Code, Falls Church, VA. 
International Conference of Building Officials, Uniform Building Code. 1997, Vol. 2, Structural 
Engineering Design Provisions. Whittier, CA. 
Ishizuka, T., and N. M. Hawkins. 1987. Effect of Bond Deterioration on the Seismic Response of 
Reinforced and Partially Prestressed Concrete and Ductile Moment Resistant Frames, SM87-2, 
University of Washington, Department of Civil Engineering. 
Jayashanker, V., and C. E. French. 1988. An Interior Moment Resistant Connection Between Precast 
Elements Subjected to Cyclic Lateral Loads, Structural Engineering Report No. 87-10, Minneapolis: 
University of Minnesota. 
Kurama, Y.C. 2002. Hybrid Post-Tensioned Precast Concrete Walls for Use in Seismic Regions. PCI 
Journal, 47, (5): 36-59. 
Mast, R. F. 1992. A Precast Concrete Frame System for Seismic Zone Four, PCI Journal 37(1):50-64. 
Nakaki, S. D., and R. E. Englekirk. 1991. PRESSS Industry Seismic Workshops: Concept 
Development, PCI Journal 36(5):54-61. 
National Fire Protection Association . 2003. Building Construction and Safety Code, NFPA 5000, 
Quincy, MA. 
Neille, D. S. 1977. Behavior of Headed Stud Connections for Precast Concrete Connections for Precast 
Concrete Panels Under Monotonic and Cycled Shear Loading, thesis submitted in partial fulfillment of 
the requirements of Doctor of Philosophy, University of British Columbia. 
New Zealand Society for Earthquake Engineering. 1991. Guidelines for the Use of Structural Precast 
Concrete in Buildings. 
Park, R., and K. J. Thompson. 1977. Cyclic Load Tests on Prestressed and Partially Prestressed 
Beam-Column Joints, PCI Journal 22(5):84-110. 
PCI Ad Hoc Committee on Precast Walls. 1997. Design for Lateral Force Resistance with Precast 
Concrete Shear Walls, PCI Journal, 42, (2): 44-65. 
Pekau, O. A., and D. Hum. 1991. Seismic Response of Friction-Jointed Precast Panel Shear Walls, 
PCI Journal 36(2):56-71. 
Powell, G., F. Filippou, V. Prakash, and S. Campbell. 1993. Analytical Platform for Precast Structural 
Systems, In Proceedings, ASCE Structures Congress '93. New York: ASCE. 
Priestley, M. J. N. 1991. Overview of PRESSS Research Program, PCI Journal 36(4):50-57. 
Priestley, M. J. N., and J. T. Tao. 1993. Seismic Response of Precast Prestressed Concrete Frames with 
Partially Debonded Tendons, PCI Journal 38(1):58-69. 
Priestley, M.J.N., S. Sritharan, J. Conley, and S. Pampanin. 1999. Preliminary Results and Conclusions 
 Concrete Structure Design Requirements 
215 
from the PRESSS Five-Story Precast Concrete Test Building, PCI Journal, 44,(6): 42-67. 
Schultz, A.E. and R.A. Magana. 1996. Seismic Behavior of Connections in Precast Concrete Walls, 
Paper SP 162-12. Mete A. Sozen Symposium, pp.273-311, ACI SP 162. American Concrete Institute, 
Farmington Hills, MI. 
Seo, S-Y., L-H Lee, and N.M. Hawkins. 1998.  The Limiting Drift and Energy Dissipation Ratio for 
Shear Walls Based on Structural Testing, Journal of the Korean Concrete Institute,10, (6): 335-343. 
Stanton, J. F., T. R. Hicks, and N. M. Hawkins. 1991. PRESSS Project 1.3: Connection Classification 
and Evaluation, PCI Journal 36(5):62-71. 
Stanton, J.F. and S.D. Nakaki, 2002. Design Guidelines for Precast Concrete Seismic Structural 
Systems-Unbonded Post-Tensioned Split Walls PRESSS Report 01/03-09, UW Report SM 02-02, 
Department of Civil Engineering, University of Washington, Seattle, WA.. 
Taylor, C.P., P.E. Cote, and J.W. Wallace.  Design of Slender Reinforced Concrete Walls with 
Openings, ACI Structural Journal, 95, (4): 420-433. 
Thompson, K. J., and R. Park. 1980. Seismic Response of Partially Prestressed Concrete, ASCE 
Journal of the Structural Division, 106(ST8):1755-1775. 
Uang, C-M., and A. Maarouf.1993. Seismic Displacement Amplification Factor in Uniform Building 
Code, SEAONC Research Bulletin Board, BB93-3, pp. B1-B2, and Displacement Amplification 
Factor for Seismic Design Provisions, Proc. Structures Congress, ASCE, Vol.1, pp. 211-216. Irvine, CA 
2 003 Commentary, Chapter 9 
216 
This page is intentionally left blank. 

217 
Appendix to Chapter 9 
UNTOPPED PRECAST DIAPHRAGMS 
Although not directly addressed in the code, untopped precast components have been used as 
diaphragms in high seismic regions. Untopped hollow-core planks with grouted joints and end chords 
have performed successfully both in earthquakes and in laboratory tests, (Elliot et al., 1992; 
Menegotto, 1994; Priestley et al., 1999). Experience has also demonstrated the unsuccessful use of 
cast-in-place concrete topping as diaphragms (Iverson and Hawkins, 1994). Where problems have 
occurred, they have not been inherently with the precast construction, but the result of a failure to 
address fundamental requirements of structural mechanics. 
This section provides conditions that are intended to ensure that diaphragms composed of precast 
components are designed with attention to the principles required for satisfactory behavior. Each 
condition addresses requirements that should be considered for all diaphragms, but which are 
particularly important in jointed construction. Specific attention should be paid to providing a 
complete load path that considers force transfer across all joints and connections. 
A9.2 DESIGN REQUIREMENTS 
A9.2.1 Configuration. Out-of-plane offsets in the vertical elements of the seismic-force-resisting 
system place particularly high demands on the diaphragm in providing a continuous load path. 
Untopped precast diaphragms are not suitable for this condition. It must be recognized that the 
demand on diaphragms in buildings with these plan irregularities requires special attention. In 
accordance with Sec. 4.6.3.2 the design force for the diaphragm should be increased by at least 25 
percent when such irregularities are present in structures assigned to Seismic Design Category D, E, or 
F. 
A9.2.2 Diaphragm demand. Following the principle that the diaphragm is not generally an 
appropriate location for inelastic behavior and, in particular, for untopped precast diaphragms, specific 
direction is provided that elastic models should be used for diaphragm analysis. Connections are 
subject to a combination of load effects (Fleischman et al., 1998). The distribution of loads may 
change after yielding, and therefore the design of the diaphragm should avoid yielding. 
Since the diaphragm is not generally an appropriate location for inelastic behavior, it should be 
designed to a level of strength that is intended to ensure that the ductility and yield strength of the 
seismic-force-resisting system can be mobilized before the diaphragm yields. While research 
(Fleischman et al., 1998) suggests that the diaphragm demand will not exceed twice the equivalent 
lateral forces used for the vertical system design, Table 4.3-1 prescribes an overstrength factor, O0
, 
and Sec. 4.3.3 prescribes a redundancy factor, ., for the systems that should be used. If an analysis of 
the probable strength of the seismic-force-resisting system is made to determine a lower demand on 
the diaphragm, the design force used should still be sufficient to attempt to ensure that the diaphragm 
remains elastic. For that reason a 1.25 factor is specified. 
A9.2.3 Mechanical connections. Although the design procedures prescribed in these sections are 
intended to ensure elastic behavior at the level of the code design forces, it is recognized that 
catastrophic events may exceed code requirements. Under such circumstances, it is important that the 
connections possess ductility under reversed cyclic loading. The intent, in these sections, is for the 
connection capacity to be limited by steel yielding of the connector and not by brittle concrete failure 
or weld fracture. 
Substantiating experimental evidence to demonstrate through testing and evaluation that mechanical 
connections satisfy the principles specified in ACI T1.1-01 and ATC-24, and can develop the required 
capacity and ductility, should meet the following criteria: 
2 003 Commentary, Appendix to Chapter 9 
218 
Test Procedures: 
1. Prior to testing, a design procedure should have been developed for prototype connections having 
the generic form that is to be tested for acceptance. 
2. That design procedure should be used to proportion the test specimens. 
3. Specimens should not be less than two-thirds scale. 
4. Test specimens should be subject to a sequence of reversing cycles having increasing limiting 
displacements. 
5. Three fully reversed cycles should be applied at each limiting displacement. 
6. The maximum load for the first sequence of three cycles should be 75 percent of the calculated 
nominal strength of the connection, En. 
7. The stiffness of the connection should be defined as 75 percent of the calculated nominal strength 
of the connection divided by the corresponding measured displacement, dm. 
8. Subsequent to the first sequence of three cycles, limiting displacements should be incremented by 
values not less than 1.0, and not more than 1.25 times dm. 
Acceptance Criteria: 
1. The connection should develop a strength, Emax, greater than its calculated nominal strength, En. 
2. The strength, Emax, should be developed at a displacement not greater than 3dm. 
3. For cycling between limiting displacements not less than 3dm, the peak force for the third loading 
cycle for a given loading direction should not be less than 0.8 Emax for the same loading direction. 
Results of reversed cyclic loading tests on typical connections are reported in Spencer (1986) and 
Pincheira et al. (1998). 
A9.2.4 Cast-in-place strips. Successful designs may include a combination of untopped precast 
components with areas of concrete topping in locations of high force demand or concentration. Such 
topping can allow for continuity of reinforcement across joints. For such designs, the requirements for 
topping slab diaphragms apply to the topped portions. 
A9.2.5 Deformation compatibility. An important element in the Provisions is attention to 
deformation compatibility requirements. Reduction in effective shear and flexural stiffness for the 
diaphragm is appropriate in evaluating the overall effects of drift on elements that are not part of the 
seismic-force-resisting system. This approach should encourage the use of more vertical elements to 
achieve shorter spans in the diaphragm and result in improved system redundancy and diaphragm 
continuity. Redundancy will also improve the overall behavior should any part of the diaphragm yield 
in a catastrophic event. 
REFERENCES 
Elliot, K. S., G. Davies, and W. Omar. 1992. Experimental and Theoretical Investigation of Precast 
Concrete Hollow-cored Slabs Used as Horizontal Floor Diaphragms. The Structural Engineer, 70 
(10) 175-187. 
Fleischman, R. B., R. Sause, S. Pessiki, and A. B. Rhodes. 1998. Seismic Behavior of Precast 
Parking Structure Diaphragms. PCI Journal, 43 (1): 38-53. 
Iverson, J. K., and N. M. Hawkins. 1994. Performance of Precast/Prestressed Concrete Building 
Structures During the Northridge Earthquake, PCI Journal 39(2): 38-55. 
Menegotto, M. 1994. Seismic Diaphragm Behavior of Untopped Hollow-Core Floors, 
Proceedings, F.I.P. Congress, Washington, D.C. 
Untopped Precast Diaphragms 
219 
Pincheira, J. A., M. G. Oliva, and F. I. Kusumo-Rahardjo. 1998. Tests on Double Tee Flange 
Connectors Subject to Monotonic and Cyclic Loading, PCI Journal, 43 (3): 82-96. 
Priestley, M. J. N., S. Sritharan, J. R. Conley, and S. Pampanin. 1999. Preliminary Results and 
Conclusions from the PRESSS Five-Story Precast Concrete Test Building. PCI Journal, 44 (6). 
Spencer, R. 1986. Earthquake Resistant Connections for Low Rise Precast Concrete Buildings, 
Proceedings, U.S.-Japan Seminar on Precast Concrete Construction in Seismic Zones, Vol. 1, Japan 
Concrete Institute. 
2 003 Commentary, Appendix to Chapter 9 
220 
This page is intentionally left blank. 

 223 
Chapter 11 Commentary 
MASONRY STRUCTURE DESIGN REQUIREMENTS 
11.1.2 References. The main concern of Chapter 11 is the proper detailing of masonry construction for 
earthquake resistance. The bulk of the detailing requirements in this chapter are contained in ACI 
530/ASCE 5/TMS 402. The commentary for ACI 530/ASCE 5/TMS 402 contains a valuable discussion 
of the rationale behind detailing requirements that is not repeated here. 
11.2.1.5.1 Shear keys. Shear keys provide resistance to the movement of shear walls when yielding of 
the reinforcing steel occurs. This phenomenon was observed in tests by Klingner. (Leiva and Klingner 
1991). There has been no field verification of shear wall movement under seismic events. The shear 
key requirements are based on judgment and sizes are based on current construction procedures. 
11.2.2.3 Article 1.3 permits the use of structural clay wall-tile meeting the requirements of ASTM C 34. 
At the time of publication, it was felt that the existing detailing requirements for masonry elements did 
not adequately address the brittle nature of clay wall-tile units. 
11.2.2.11 The nominal shear strength of coupling beams must be equal to the shear caused by 
development of a full yield hinge at each end of the coupling beams. This nominal shear strength is 
estimated by dividing the sum of the calculated yield moment capacity of each end of the coupling 
beams, M1 and M2, by the clear span length, L. 
A coupling beam may consist of a masonry beam and a part of the reinforced concrete floor system. 
Reinforcement in the floor system parallel to the coupling beam should be considered as a part of the 
coupling beam reinforcement. The limit of the minimum width of floor that should be used is six times 
the floor slab thickness. This quantity of reinforcement may exceed the limits of Sec. 3.2.3.5 but should 
be used for the computation of the normal shear strength. 
11.2.2.12 The theory used for design of beams has a limited applicability to deep beams. Shear warping 
of the cross section and a combination of diagonal tension stress and flexural tension stress in the body 
of the deep beam requires that deep beam theory be used for design of members that exceed the 
specified limits of span to depth ratio. Analysis of wall sections that are used as beams generally will 
result in a distribution of tensile stress that requires the lower one-half of the beam section to have 
uniformly distributed reinforcement. The uniform distribution of reinforcement resists tensile stress 
caused by shear as well as flexural moment. 
The flexural reinforcement for deep beams must meet or exceed the minimum flexural reinforcement 
ratio of Sec. 3.2.4.3.2. Additionally, horizontal and vertical reinforcement must be distributed 
throughout the length and depth of deep beams and must provide reinforcement ratios of at least 
0.0007bd. Distributed flexural reinforcement may be included in the calculations of the minimum 
distributed reinforcement ratios. 
11.2.2.13 Corrugated sheet metal ties are prohibited from use in Seismic Design Categories E and F due 
to their decreased capacity in transferring loads. 
11.2.2.14 Masonry pryout refers to a failure mode of a shear anchor in which the embedded end of the 
anchor moves opposite to the direction of applied shear, prying out a roughly semi-conical body of 
masonry (concrete, as applicable) behind the anchor. It is not the same as a breakout, which refers to 
a failure mode of a shear anchor in which a body of masonry (or concrete, as applicable) is broken off 
between the anchor and a free edge, in the direction of applied shear. 

2 003 Commentary, Chapter 11 
224 
11.4 GLASS-UNIT MASONRY AND MASONRY VENEER 
Chapters 11 and 12 of ACI 530-95/ASCE 5-95/TMS 402-95 were introduced into the 1997 Provisions 
to address design of glass-unit masonry and masonry veneer. Direct reference is made to these chapters 
for design requirements. Investigations of seismic performance have shown that architectural 
components meeting these requirements perform well (Jalil, Kelm, and Klingner, 1992; and Klingner, 
1994). 
11.5 PRESTRESSED MASONRY 
Allowable stress provisions are set forth in MSJC Chapter 4. There are no strength design provisions for 
prestressed masonry. There is a paucity of data on the cyclic testing of prestressed shear walls. There is 
only one published report of cyclic testing of prestressed shear walls in-plane using a testing protocol 
similar to the sequential phased displacement method used in the TCCMaR program. This report 
considers specimens both partially and fully grouted using only prestressed bar reinforcing. There is no 
published in-plane cyclic test data using prestressed strand, nor any published data using prestressed 
reinforcing in combination with mild steel reinforcing. There is some additional unpublished data on inplane 
testing of prestressed masonry shear walls using prestressed bars only. 
The data shows that solid grouted prestressed masonry shear walls subjected to in-plane cyclic 
displacements perform as an essentially elastic system with stiffness degradation in each cycle. Little 
energy is dissipated in the hysterisis loops. Although reasonably large displacements can be reached, 
there is essentially no ductile behavior. The data on partially grouted walls is sparse and shows inability 
to reach large displacement before failure. The data shows that MSJC Eq. 3-21 provides a reasonable 
estimate for the shear capacity for solid grouted walls. 
The TCCMaR research showed that the ductility of a masonry wall loaded in-plane was highly 
dependent on the level of axial load and the amount of reinforcing. Ductile behavior declines 
significantly at axial loads in excess of 100 psi; ductile behavior also declines significantly when the 
reinforcement ratio is high. The addition of prestressing to a wall with mild steel reinforcing will 
decrease the ductility. 
Because of the limited data and the potential for non-ductile, prestressed masonry shear walls are 
restricted to Seismic Design Categories A and B and the R factor is set at 1. As more research 
becomes available, these restrictions could be eased. 
11.6 ANCHORING TO MASONRY 
This section covers cast-in-place headed anchor bolts and bent-bar anchors (J- or L-bolts) in grout. 
General background information on this topic is given in CEB, 1995. 
The tensile capacity of a headed anchor bolt is governed by yield and fracture of the anchor steel or by 
breakout of a roughly conical volume of masonry starting at the anchor head and having a fracture 
surface oriented at 45 degrees to the masonry surface. Steel capacity is calculated using the effective 
tensile stress area of the anchor (that is, including the reduction in area of the anchor shank due to 
threads). Masonry breakout capacity is calculated using expressions adapted from concrete design, 
which use a simplified design model based on a stress of 4 fm' uniformly distributed over the area of 
that right circular cone, projected onto the surface of the masonry. Reductions in breakout capacity due 
to nearby edges or adjacent anchors are computed in terms of reductions in those projected areas 
(Brown and Whitlock, 1983). 
The tensile capacity of a bent-bar anchor bolt (J- or L-bolt) is governed by yield and fracture of the 
anchor steel, by tensile cone breakout of the masonry, or by straightening and pullout of the anchor from 
the masonry. Capacities corresponding to the first two failure modes are calculated as for headed 
anchor bolts. Pullout capacity is calculated as proposed by Shaikh (1996). Possible contributions to 
tensile pullout capacity due to friction are neglected. 
 Masonry Structure Design Requirements 
225 
The tensile breakout capacity of a headed anchor is usually much greater than the pullout capacity of a 
J- or L-bolt. The designer is encouraged to use headed anchors when anchor tensile capacity is critical. 
The shear capacity of a headed or a bent-bar anchor bolt is governed by yield and fracture of the anchor 
steel or by masonry shear breakout. Steel capacity is calculated using the effective tensile stress area 
(that is, threads are conservatively assumed to lie in the critical shear plane). Shear breakout capacity is 
calculated as proposed by Brown and Whitlock, 1983. 
Under static shear loading, bent-bar anchor bolts (J- or L-bolts) do not exhibit straightening and pullout. 
Under reversed cyclic shear, however, available research suggests that straightening and pullout may 
occur. Headed anchor bolts are recommended for such applications (Malik et al., 1982). 
REFERENCES 
Abboud, B. E., X. Lu, and F. C. Schmitt. 1993. An Evaluation of Three Current Deflection Methods 
for Predicting the Lateral Deflection of Masonry Walls. In Proceedings of the Sixth North American 
Masonry Conference, Philadelphia, Pennsylvania, pp.73-85. 
Abboud, B. E., and A. A. Hamid. 1987. A Comparative Study of the Flexural Behavior of Reinforced 
Block Masonry and Reinforced Concrete Using Small Scale Model Walls. In Proceedings of the 
Fourth North American Masonry Conference, Los Angeles, California. 
Abboud, B. E., A. A. Hamid, and H. G. Harris. 1990. Small Scale Modeling of Concrete Block 
Masonry Structures, ACI Structural Journal 87(2):145-155. 
Abboud, B. E. 1987. The Use of Small Scale Direct Models for Concrete Block Masonry Assemblages 
and Slender Reinforced Walls Under Out-of-Plane Loads, Doctoral Thesis, Drexel University, 
Philadelphia, Pennsylvania. 
Blondet, J. M., R. L. Mayes, T. E. Kelley, R. R. Villablanca, and R. E. Klingner. 1989. Performance of 
Engineered Masonry in the Chilean Earthquake of March 3, 1985: Implications for U.S. Design 
Practice. Austin: Phil Ferguson Structural Engineering Laboratory, University of Texas. 
Brown, R. H., and A. R. Whitlock. 1983. Strength of Anchor Bolts in Grouted Masonry, ASCE 
Journal of Structural Engineering 109(6). 
CEB. 1995. Design of Fastenings in Concrete (Draft CEB Guide, Parts 1 to 3) and Fastenings for 
Seismic Retrofitting (State-of-the-Art Report on Design and Application), Task Group 3.5 
(Embedments), Euro-International Concrete Committee (CEB), CEB Bulletin D=Information 226. 
Comite Euro-International du Beton. 
Chen, S. W., P. A. Hidalgo, R. L. Mayes, R. W. Clough, and H. D. McNiven. 1978. Cyclic Loading 
Tests of Masonry Single Piers, Vol. 2, EERC Report UCB/EERC-78/27. Berkeley: University of 
California Earthquake Engineering Research Center. 
Fattal, S. G. 1993a. Strength of Partially Grouted Masonry Shear Walls, Report NIST 5147. 
Gaithersburg, Maryland: National Institute of Standards and Technology. 
Fattal, S. G. 1993b. The Effect of Critical Parameters on the Behavior of Partially Grouted Masonry 
Shear Walls Under Lateral Loads, NIST Report NISTIR 5116. Gaithersburg, Maryland: National 
Institute of Standards and Technology. 
Fattal, S. G., and D. R. Todd. 1991. Ultimate Strength of Masonry walls: Prediction vs. Test Results, 
NIST Report NISTIR 4633. Gaithersburg, Maryland: National Institute of Standards and Technology. 
Hamid, A. A., M. K. Hatem, H. G. Harris, and B. E. Abboud. 1990. Hysteretic Response and 
Ductility of Reinforced Concrete Masonry Walls Under Out-of-Plane Loading. In Proceedings of the 
Fifth North American Conference, Urbana, Illinois, pp. 397-410. 
2 003 Commentary, Chapter 11 
226 
Hamid, A. A., B. E. Abboud, M. W. Farah, M. K. Hatem, and H. G. Harris. 1989. Response of 
Reinforced Block Masonry Walls to Out-of-Plane Static Loads. In U.S.-Japan Coordinated Program 
For Masonry Building Research, Report 3.2(a). Drexel University. 
He, L., and N. Priestley. 1992. Seismic Behavior of Flanged Masonry Shear Walls - Final Report, 
TCCMaR Report 4.1-2. 
Hidalgo, P. A., R. L. Mayes, H. D. McNiven, and R. W. Clough. 1978. Cyclic Loading Tests of 
Masonry Single Piers, Vol. 1 and 3, EERC Report UCB/EERC-78/27 and 79/12. Berkeley, University 
of California Earthquake Engineering Research Center. 
Horton, R. T., and M. K. Tadros. 1990. Deflection of Reinforced Masonry Members, ACI Structural 
Journal 87(4):73-85. 
Jail, I., W. Kelm, and R. E. Klingner. 1992. Performance of Masonry and Masonry Veneer Buildings 
in the Loma Prieta Earthquake, PMSEL Report 92-1. Austin: University of Texas, Department of 
Civil Engineering. 
Klingner, R. E. 1994. Performance of Masonry Structures in the Northridge, California, Earthquake of 
January 17, 1991. Boulder, Colorado: The Masonry Society. 
Leiva, G., and R. Klingner. 1991. In-Plane Seismic Resistance of Two-story Concrete Masonry Shear 
Walls with Openings, TCCMaR Report 3.1(c)-2. 
Limin, H., and N. Priestley. 1988. Seismic Behavior of Flanged Masonry Shear Walls, TCCMaR 
Report 4.1-1. 
Malik, J. B., J. A. Mendonca, and R. E. Klingner. 1982. Effect of Reinforcing Details on the Shear 
Resistance of Short Anchor Bolts Under Reversed Cyclic Loading, Journal of the American Concrete 
Institute, Proceedings, 79, (1): 3-11. 
Matsumura, A. 1987. Shear Strength of Reinforced Hollow Unit Masonry Walls. In Proceedings of 
the 4th North American Masonry Conference, Los Angeles, California. 
Mayes, R. L., Y. Omote, and R. W. Clough. 1976. Cyclic Shear Tests of Masonry Piers, Vol. 1, Report 
76-8. Berkeley: University of California Earthquake Engineering Research Center. 
Merryman, K., G. Leiva, N. Antrobus, and R. Klingner. 1989. In-Plane Seismic Resistance of Twostory 
Concrete Masonry Coupled Shear Walls, TCCMaR Report 3.1(c)-1. 
Okamoto, S., Y. Yamazaki, T. Kaminosono, M. Teshigawara, and H. Hirashi. 1987. Seismic Capacity 
of Reinforced Masonry Walls and Beams in Wind and Seismic Effects. In Proceedings of the 18th 
joint Meeting, U.S. - Japan Panel on Wind and Seismic Effects, ed. by N. J. Raufaste, Report NBSIR 
87-3540. Gaithersburg, Maryland: National Institute of Standards and Technology. 
Priestley, N., and H. Limin. 1990. Seismic Response of T-Section Masonry Shear Walls. TMS 
Journal 9(1):10-19. 
Seible, F., G. A. Hegemier, M. J. N. Priestley, G. R. Kingsley, A. Kurkchubasche, and A. Igarashi. 
1992. The U.S. - TCCMaR Five-story Full Scale Masonry Research Building Test - Preliminary Report, 
TCCMaR Report 9.2-4. 
Shaikh, A. Fattah. 1996. Design of Hooked Anchor bolts in Concrete and Masonry: Proposed Code 
Provisions and Commentary. Prepared for the National Codes and Standards Council, 1996. 
Shing, P. B., J. Noland, H. Spaeh, E. Klamerus, and M. Schuller. 1991. Response of Singlestory 
Reinforced Masonry Shear Walls to In-Plane Lateral Loads, TCCMaR Report 3.1(a)-2. 
Shing, P. B., M. Schuller, and V. S. Hoskere. 1990. In-Plane Resistance of Masonry Shear Walls, 
ASCE Journal of Structural Engineering 116(3). 
 Masonry Structure Design Requirements 
227 
Sveinsson, B. I., H. D. McNiven, and H. Sucuoglu. 1985. Cyclic Shear Tests of Masonry Piers, Vol. 4, 
Report UCB/EERC 85-15. Berkeley: University of California Earthquake Engineering Research 
Center. 
Wolde-Tinsae, A. M., R. H. Atkinson, and A. A. Hamid. 1993. State-of-the-Art Modulus of Elasticity 
of Masonry. In Proceedings of the Sixth North American Masonry Conference. Boulder, Colorado: 
The Masonry Society. 
The following Technical Coordinating Committee for Masonry Research task reports not specifically 
cited but providing the substantiating data for the strength design criteria presented in this chapter are 
available through the Earthquake Engineering Research Center Library in Richmond, California (phone 
415-231-9403): 
Task No. Author(s) and Title 
1.1-1: Atkinson and Kingsley, Comparison of the Behavior of Clay & Concrete Masonry in 
Compression, September 1985. 151 pgs. 
1.2(a)-1: Hamid, A. A., G. F. Assis, and H. G. Harris, Material Models for Grouted Block Masonry, 
August 1988. 67 pgs. 
1.2(a)-2: Assis, G. F., A. A. Hamid, and H. G. Harris, Material Models for Grouted Block Masonry, 
August 1989. 134 pgs. 
1.2(b)-1: Young, J. M., and R. H. Brown, Compressive Stress Distribution of Grouted Hollow Clay 
Masonry Under Strain Gradient, May 1988, 170 pgs. 
1.3-1: Atkinson, R. H., An Assessment of Current Material Test Standards for Masonry Limit 
States Design Methods, June 1991. 38 pgs. 
2.1-1: Hart, G., and M. Basharkhah, Slender Wall Structural Engineering Analysis Computer 
Program (Shwall, Version 1.01), September 1987. 68 pgs. 
2.1-2: Hart, G., and M. Basharkhah, Shear Wall Structural Engineering Analysis Computer 
Program (Shwall, Version 1.01). September 1987. 75 pgs. 
2.1-3: Nakaki, D., and G. Hart, Uplifting Response of Structures Subjected to Earthquake Motions, 
August 1987. 200 pgs. 
2.1-4: Hart, G., N. Sajjad, and M. Basharkhah, Inelastic Column Analysis Computer Program 
(INCAP, Version 1.01), March 1988. 
2.1-5: Hong, W. K., G. C. Hart, and R. E. Englekirk, Force-Deflection Evaluation and Models for 
University of Colorado Flexural Walls, December 1989. 
2.1-6: Hart, G. C., J. W. Jaw, and Y. K. Low, SCM Model for University of Colorado Flexural 
Walls, December 1989. 31 pgs. 
2.1-7: Hart, G. C., N. Sajjad, and M. Basharkhah, Inelastic Masonry Flexural Shear Wall Analysis 
Computer Program, February 1990. 41 pgs. 
2.1-8: Hart, G. C., R. Englekirk, M. Srinivasan, S. C. Huang, and D. J. Drag, Seismic Performance 
Study, DPC Gymnasium, Elastic Time History Analysis Using SAP90, February 1992. 41 
pgs. 
2.1-9: Hart, G. C., R. Englekirk, M. Srinivasan, S. C. Huang, and D. J. Drag, Seismic Performance 
Study, TMS Shopping Center, Elastic Time History Analysis Using SAP90, February 1992. 
42 pgs. 
2.1-10: Hart, G. C., R. Englekirk, J. W. Jaw, M. Srinivasan, S. C. Huang, and D. J. Drag, Seismic 
Performance Study, RCJ Hotel, February 1992. 51 pgs. 
2 003 Commentary, Chapter 11 
228 
2.1-11: Hart, G. C., R. Englekirk, M. Srinivasan, S. C. Huang, and D. J. Drag, Performance Study, 
2-Story Masonry Wall-Frame Building, February 1992. 112 pgs. 
2.1-12: Hart, G. C., R. Englekirk, J. W. Jaw, M. Srinivasan, S. C. Huang, and D. J. Drag, Seismic 
Performance Study, Designed by Tentative Limit Sates Design Standard, February 1992. 
75 pgs. 
2.2-1: Ewing, R. D., A. El-Mustapha, and J. Kariotis, FEM/I - A Finite Element Computer 
Program for the Nonlinear Static Analysis of Reinforced Masonry Building Components, 
December 1987 (Revised June 1990). 124 pgs. 
2.2-2: Ewing, R. D., Parametric Studies on Reinforced Masonry Shear Walls Using FEM/I, A 
Nonlinear Finite Element Analysis Program, March 1992. 
2.2-3: Ewing, R. D., Finite Element Analysis of Reinforced Masonry Building Components 
Designed by a Tentative Masonry Limit States Design Standard, March 1992. 48 pgs. 
2.3-1: Ewing, R., J. Kariotis, and A. El-Mustapha, LPM/I, A Computer Program for the 
Nonlinear, Dynamic Analysis of Lumped Parameter Models, August, 1987. 200 pgs. 
2.3-2: Kariotis, J., A. El-Mustapha, and R. Ewing, Influence of Foundation Model on the Uplifting 
of Structures, July 1988. 50 pgs. 
2.3-3: Kariotis, J., A. Rahman, and A. El-Mustapha, Investigation of Current Seismic Design 
Provisions for Reinforced Masonry Shear Walls, January 1990. 48 pgs. 
2.3-4: Kariotis, J., A. Rahman, O. Waqfi, and R. Ewing, Version 1.03 LPM/I - A Computer 
Program for the Nonlinear, Dynamic Analysis of Lumped Parameter Models, February 
1992. 227 pgs. 
2.3-5: Kariotis, J., O. Waqfi, and R. Ewing, R., A Computer Program Using Beam Elements for 
the Nonlinear, Dynamic Analysis of Lumped Parameter Models, February 1992. 96 pgs. 
2.3-6: Kariotis, J., and O. Waqfi, Comparison of the Dynamic Response of a Damped MDOF 
Nonlinear Beam Model with an Equivalent SDOF Hysteretic Model, April 1992. 88 pgs. 
2.3-7: Kariotis, J., and O. Waqfi, Recommended Procedure for Calculation of the Balanced 
Reinforcement Ratio, February 1992. 73 pgs. 
2.4(b)-1: Button, M. R., and R. L. Mayes, Out-of-Plane Seismic Response of Reinforced Masonry 
Walls: Correlation of Full-Scale Test and Analytical Model Results, March 1991. 65 pgs. 
3.1(a)-1: Scrivener, J., Summary of Findings of Cyclic Tests on Masonry Piers, June 1986. 7 pgs. 
3.1(a)-2: Shing, P. B., J. Noland, H. Spaeh, E. Klamerus, and M. Schuller, Response of Single-Story 
Reinforced Masonry Shear Walls to In-Plane Lateral Loads, January 1991. 136 pgs. 
3.1(b)-1: Seible, F., and H. LaRovere, Summary of Pseudo Dynamic Testing, February 1987. 46 pgs. 
3.1(b)-2: Igarashi, A., F. Seible, and G. Hegemier, Development of the Generated Sequential 
Displacement Procedure and the Simulated Seismic Testing of the TCCMaR Three-Story In- 
Plane Walls, June 1993. 
3.1(c)-1: Merryman, K., G. Leiva, B. Antrobus, and R. Klingner, In-Plane Seismic Resistance of 
Two-Story Concrete Masonry Coupled Shear Walls, September 1989. 176 pgs. 
3.1(c)-2: Leiva, G., and R. Klingner, In-plane Seismic Resistance of Two-story Concrete Masonry 
Shear Walls with Openings, August 1991. 326 pgs. 
3.2(a)-1: Hamid, A., B. Abboud, M. Farah, K. Hatem, and H. Harris, Response of Reinforced Block 
Masonry Walls to Out-of-Plane Static Loads, September 1989. 120 pgs. 
 Masonry Structure Design Requirements 
229 
3.2(b)-1: Agbabian, M., S. Adham, S. Masri, V. Avanessian, and V. Traina, Out-of-Plane Dynamic 
Testing of Concrete Masonry Walls, Vol. 1 and 2, July 1989. 220 pgs. 
3.2(b)-2: Blondet, M., and R. L. Mayes, The Transverse Response of Clay Masonry Walls Subjected 
to Strong Motion Earthquakes, Vol. 1: General Information, April 1991. 172 pgs. 
3.2(b)-2: Blondet, M., and R. L. Mayes, The Transverse Response of Clay Masonry Walls Subjected 
to Strong Motion Earthquakes, Vol. 2: Walls No. 4 and 6 (Group 1), April 1991. 267 pgs. 
3.2(b)-2: Blondet, M., and R. L. Mayes, The Transverse Response of Clay Masonry Walls Subjected 
to Strong Motion Earthquakes, Vol. 3: Walls No. 8, 9, 10 and 11 (Group 2), April 1991. 
310 pgs. 
3.2(b)-2: Blondet, M., and R. L. Mayes, The Transverse Response of Clay Masonry Walls Subjected 
to Strong Motion Earthquakes, Vol. 4: Walls No. 3, 5, and 7 (Group 3), April 1991. 256 
pgs. 
4.1-1: He, L., and M. J. N. Priestley, Seismic Behavior of Flanged Masonry Shear Walls, May 
1988. 119 pgs. 
4.1-2: He, L., and M. J. N. Priestley, Seismic Behavior of Flanged Masonry Shear Walls - Final 
Report, November 1992. 279 pgs. 
4.2-1: Hegemier, G., and H. Murakami, On the Behavior of Floor-to-Wall Intersections in 
Concrete Masonry Construction, Part I: Experimental. 
4.2-2: Hegemier, G., and H. Murakami, On the Behavior of Floor-to-Wall Intersections in 
Concrete Masonry Construction, Part II: Theoretical. 
5.1-1: Porter, M., and A. Sabri, Plank Diaphragm Characteristics, July 1990. 226 pgs. 
5.2-1: Porter, M., F. Yeomans, and A. Johns, Assembly of Existing Diaphragm Data, July 1990. 
142 pgs. 
6.2-1: Scrivener, J., Bond of Reinforcement in Grouted Hollow-Unit Masonry: A State-of-the-Art, 
June 1986. 53 pgs. 
6.2-2: Soric, Z., and L. Tulin, Bond Splices in Reinforced Masonry, August 1987. 296 pgs. 
7.1-1: Paulson, T., and D. Abrams, Measured Inelastic Response of Reinforced Masonry Building 
structures to Earthquake Motions, October 1990. 294 pgs. 
8.1-1: Hart, G., A Limit State Design Method for Reinforced Masonry, June 1988. 
8.1-2: Hart, G., Expected Value Design in the Context of a Limit Sate Design Methodology, 
February 1990. 
8.2-1: Hart, G., and G. T. Zorapapel, Reliability of Concrete Masonry Wall Structures, December 
1991. 229 pgs. 
8.2-2: Hart, G., and N. Sajjad, Confinement in Concrete Masonry, December 1990. 
8.2-3: Hart, G., and J. Jang, Seismic Performance of Masonry Wall Frames, December 1991. 
9.1-1: Kariotis, J. C., and A. W. Johnson, Design of Reinforced Masonry Research Building, 
September 1987. 42 pgs. 
9.1-2: Kariotis, J. C., and O. M. Waqfi, Trial Designs Made in Accordance with Tentative Limit 
States Design Standards for Reinforced Masonry Buildings, February 1992. 184 pgs. 
9.2-1: Seible, F., Report on Large Structures Testing Facilities in Japan, September 1985. 120 
pgs. 
2 003 Commentary, Chapter 11 
230 
9.2-2: Seible, F., Design and Construction of the Charles Lee Powell Structural Systems 
Laboratory, November 1986. 65 pgs. 
9.2-3: Seible, F., The Japanese Five-story Full Scale Reinforced Masonry Building Test, January 
1988. 100 pgs. 
9.2-4: Seible, F., G. A. Hegemier, M. J. N. Priestley, G. R. Kingsley, A. Kurkchubasche, and A. 
Igarashi, The U.S. - TCCMaR Five-story Full Scale Masonry Research Building Test - 
Preliminary Report, October 1992. 58 pgs. 
11.1-1: TCCMaR, Summary Report: U.S. Coordinated Program for Masonry Building Research, 
September 1985 to August 1986. 190 pgs. 
11.1-2: TCCMaR, Status Report: U.S. Coordinated Program for Masonry Building Research, 
November 1988. 170 pgs. 

 231 
Chapter 12 Commentary 
WOOD STRUCTURE DESIGN REQUIREMENTS 
12.1 GENERAL 
12.1.2 References. Wood construction practices have not been codified in a form that is standard 
throughout the country. The 2003 Provisions incorporates by reference the AF&PA ASD/LRFD 
Supplement, Special Design Provisions for Wind and Seismic (SDPWS) and the 2003 International 
Residental Code (IRC). Many wood frame structures are a combination of engineered wood and 
conventional light-frame construction. Wood also is used in combination with other materials 
(American Institute of Timber Construction, 1985; Breyer, 1993; Faherty and Williamson, 1989; 
Hoyle and Woeste, 1989; Somayaji, 1992; Stalnaker and Harris, 1989). The requirements of the 
model building codes were used as a resource in developing the requirements introduced in the 1991 
Provisions and further modified since then. The general requirements of Chapter 12 cover construction 
practices necessary to provide a performance level of seismic resistance consistent with the purposes 
stated in Chapter 1. These requirements also may be related to gravity load capacity and wind force 
resistance which is a natural outgrowth of any design procedure. For the 2003 Provisions, the 
reference documents continue to be grouped according to their primary focus into three subsections: 
Sec. 12.1.2.1, Engineered Wood Construction; Sec. 12.1.2.2, Conventional Construction; and Sec. 
12.1.2.3, Materials Standards. 
12.2 DESIGN METHODS 
Prior to the publication of AF&PA/ASCE 16, typical design of wood frame structures followed the 
American Forest and Paper Association (AF&PA) National Design Specification for Wood 
Construction (NDS) (AF&PA, 1991). The NDS is based on allowable stresses and implied factors 
of safety. However, the design procedure provided by the Provisions was developed on the premise of 
the resistance capacity of members and connections at the yield level (ASCE, 1988; Canadian Wood 
Council, 1990 and 1991; Keenan, 1986). In order to accommodate this difference in philosophy, the 
1994 and prior editions of the Provisions made adjustments to the tabulated allowable stresses in the 
reference documents. 
With the completion of the Load and Resistance Factor Standard for Engineered Wood Construction 
(AF&PA/ASCE, 1995), the modifications and use of an allowable stress based standard was no 
longer necessary. Therefore, the 1997 Provisions included the LRFD standard by reference 
(AF&PA/ASCE 16) and used it as the primary design procedure for engineered wood construction. 
The use of AF&PA/ASCE 16 continues in the 2003 Provisions. 
Conventional light-frame construction, a prescriptive method of constructing wood structures, is 
allowed for some design categories. These structures must be constructed according to the 
requirements set forth in Sec. 12.4 and applicable reference documents. If the construction deviates 
from these prescriptive requirements, the engineered design requirements of Sec. 12.2 and 12.3 and 
AF&PA/ASCE 16 must be followed. If a structure that is classified as conventional construction 
contains some structural elements that do not meet the requirements of conventional construction, the 
elements in question can be engineered without changing the rest of the structure to engineered 
construction. The extent of design to be provided must be determined by the responsible registered 
design professional; however, the minimum acceptable extent is often taken to be force transfer into 
the element, design of the element, and force transfer out of the element. This does not apply to a 
structure that is principally an engineered structure with minor elements that could be considered 
conventional. When more than one braced wall line or diaphragm in any area of a conventional 
residence requires design, the nature of the construction may have changed, and engineered design 
2003 Commentary, Chapter 12 
232 
might be appropriate for the entire seismic-force-resisting system. The absence of a ceiling diaphragm 
may also create a configuration that is non-conventional. The requirement for engineering portions of 
a conventional construction structure to maintain lateral-force resistance and stiffness is added to 
provide displacement compatibility. 
Alternate strength of members and connections. It remains the intent of the Provisions that load 
and resistance factor design be used. When allowable stress design is to be used, however, the 
factored resistance of members and connections subjected to seismic forces acting alone or in 
combination with other prescribed loads shall be determined using a capacity reduction factor, f, times 
2.16 times the allowable stresses permitted in the National Design Specification for Wood 
Construction (NDS) and supplements (AF&PA, 1991). The allowable stresses used shall not include a 
duration of load factor, CD. The value of the capacity reduction factor, f, shall be as follows: 
Wood members 
In flexure f = 1.00 
In compression f = 0.90 
In tension f = 1.00 
In shear and torsion f = 1.00 
Connectors 
Anchor bolts, bolts, lag bolts, nails, screws, etc. f = 0.85 
Bolts in single shear in members of a 
seismic-force-resisting system f = 0.40 
These soft conversions from allowable stress design values to load and resistance factor design 
values first appeared in Sec. 9.2 in the 1994 Provisions. An alternative method of calculating soft 
conversions is provided in ASTM D 5457-93. The reader is cautioned, however, that the loads and 
load combinations to be used for conversion are not specified so it is incumbent upon the user to 
determine appropriate conversion values. Wood frame structures assigned to Seismic Design 
Category A, other than one- and two-family dwellings, must comply with Sec. 12.4 or if engineered 
need only comply with the reference documents and Sec. 1.5. Exceptions addressing one- and twofamily 
detached dwellings appear in Sec. 
12.2.1 Seismic Design Categories B, C, and D. Seismic Design Categories B, C, and D were 
combined in the 1997 Provisions. At the same time, subsections on material limitations and 
anchorage requirements were moved. This was based on the philosophy that detailing requirements 
should vary based on R value rather than seismic design category. 
Structures assigned to Seismic Design Categories B, C, and D are required to meet the minimum 
construction requirements of Sec. 12.4 (Sherwood and Stroh, 1989) or must be engineered using 
standard design methods and principles of mechanics. Conventional light-frame construction 
requirements were modified in the 1991 Provisions to limit the spacing between braced wall lines 
based on calculated capacities to resist the loads and forces imposed. 
Engineered structures assigned to Seismic Design Categories B, C, and D are required to conform to 
the provisions of Sec. 12.2 and 12.3. Included in these sections are general design limitations, limits 
on wood resisting forces contributed by concrete or masonry, shear wall and diaphragm aspect ratio 
limitations, and requirements for distribution of shear to vertical resisting elements. 
Wood Structure Design Requirements 
233 
12.2.2 Seismic Design Categories E and F. If the provisions of Chapter 12 apply, Seismic Design 
Category E and F structures require an engineered design. Conventional construction is not 
considered rigorous enough for structures expected to be functional following a major seismic event. 
For Seismic Design Category E and F structures, close attention to load path and detailing is required. 
Structures assigned to Seismic Design Category E and F require blocked diaphragms. Structural-use 
panels must be applied directly to the framing members; the use of gypsum wallboard between the 
structural-use panels and the framing members is prohibited because of the poor performance of nails 
in gypsum. Restrictions on allowable shear values for structural-use shear panels when used in 
conjunction with concrete and masonry walls are intended to provide for deformation compatibility of 
the different materials. 
12.2.3.1 Discussion of cyclic test protocol is included in ATC (1995), Dolan (1996), and Rose (1996). 
12.2.3.2 and 12.2.3.7 The mid-span deflection of a simple-span, blocked wood structural panel 
diaphragm uniformly nailed throughout may be calculated by use of the following formula: 
5 3 0.188 ( ) 
8 4 2 c 
n 
vL vL Le X 
bEA Gt b 
S . 
. = + + + 
where: 
. = the calculated deflection, in. (mm). 
v = maximum shear due to factored design loads in the direction under consideration, lb/ft 
(kN/m). 
L = diaphragm length, ft (m). 
b = diaphragm width, ft (m). 
E = elastic modulus of chords, psi (MPa). 
A = area of chord cross-section, in.2 (mm2). 
Gt = panel rigidity through the thickness, lb/in. (N/mm). 
en = nail deformation, in. (mm). 
S (.cX) = sum of individual chord-splice slip values on both sides of the diaphragm, each 
multiplied by its distance to the nearest support, in. (mm). 
If not uniformly nailed, the constant 0.188 in the third term must be modified accordingly. See ATC 7 
(Applied Technology Council, 1981). 
This formula was developed based on engineering principles and monotonic testing. Therefore, it 
provides an estimate of diaphragm deflection due to loads applied in the factored resistance shear 
range. The effects of cyclic loading and resulting energy dissipation may alter the values for nail 
deformation in the third term, as well as chord splice effects of the fourth term, if mechanically-spliced 
wood chords are used. The formula is not applicable to partially-blocked diaphragms. 
The deflection of a blocked wood structural panel shear wall may be calculated by use of the following 
formula. 
8 3 0.75 n a 
vh vh he h d 
bEA Gt b 
. = + + + 
where: 
. = the calculated deflection, in. (mm). 
v = maximum shear due to factored design loads at the top of the wall, lb/ft (kN/m). 
2003 Commentary, Chapter 12 
234 
h = shear wall height, ft (m). 
b = shear wall width, ft (m). 
E = elastic modulus of boundary element (vertical member at shear wall boundary), psi (MPa). 
A = area of boundary element cross-section (vertical member at shear wall boundary), in.2 
(mm2). 
Gt = panel rigidity through the thickness, lb/in. (N/mm). 
en = nail deformation, in. (mm). 
da = deflection due to anchorage details ( rotation and slip at hold downs), in. (mm). 
Guidance for use of the above two equations can be found in the references. 
One stipulation is that there are no accepted rational methods for calculating deflections for 
diaphragms and shear walls that are sheathed with materials other than wood structural panel products 
fastened with nails. Therefore, if a rational method is to be used, the capacity of the fastener in the 
sheathing material must be validated by acceptable test procedures employing cyclic forces or 
displacements. Validation must include correlation between the overall stiffness and capacity 
predicted by principles of mechanics and that observed from test results. A diaphragm or shear wall 
sheathed with dissimilar materials on the two faces should be designed as a single-sided wall using the 
capacity of the stronger of the materials and ignoring the weaker of the materials. 
TABLE C12.2A 
en FASTENER SLIP EQUATIONS FOR USE IN CALCULATING DIAPHRAGM 
AND SHEAR WALL DEFLECTION DUE TO FASTENER SLIP 
Fastener Slip, en (in.)1 
Fastener 
Minimum 
Penetration 
(in.) 
Maximum Fastener 
Loads - Vn 
(lb/fastener) 
Fabricated w/green 
(>19% m.c.) 
lumber 
Fabricated 
w/dry (< 19% 
m.c.) lumber 
6d common nail 1-1/4 180 (Vn/434)2.314 (Vn/456)3.144 
8d common nail 1-3/8 220 (Vn/857)1.869 (Vn/616)3.018 
10d common nail 1-1/2 260 (Vn/977)1.894 (Vn/769)3.276 
14-ga staple 1 to <2 140 (Vn/902)1.464 (Vn/596)1.999 
14-ga staple >2 170 (Vn/674)1.873 (Vn/461)2.776 
For SI: 1 inch = 25.4 mm, 1 pound = 4.448 N. 
1.Values apply to plywood and OSB fastened to lumber with a specific gravity of 0.50 or greater except that the 
slip shall be increased by 20 percent when plywood is not Structural I. 

Wood Structure Design Requirements 
235 
TABLE C12.2B 
VALUES OF Gt FOR USE IN CALCULATING DEFLECTION OF 
WOOD STRUCTURAL PANEL DIAPHRAGMS AND SHEAR WALLS 

2 003 Commentary, Chapter 12 
236 
VALUES OF Gt (lb/in. panel depth or width) 
STRUCTURAL I OTHER 
PANEL 
TYPE 
Thickness 
(in.) All Plywood Grades Marine All Other 
Plywood 
Grades 
 31,000 31,000 24,000 
11/32 33,000 33,000 25,500 
3/8 34,000 34,000 26,000 
15/32 49,500 49,500 38,000 
 50,000 50,000 38,500 
19/32 63,500 63,500 49,000 
5/8 64,500 64,500 49,500 
23/32 65,500 65,500 50,500 
 66,500 66,500 51,000 
7/8 68,500 68,500 52,500 
1 95,500 95,500 73,500 
Sanded 
Plywood 
1-1/8 97,500 97,500 75,000 
For SI: 1 inch = 25.4 mm, 1 pound/inch of panel depth or width = 0.1751 N/mm. 
1. Applies to plywood with 5 or more layers; for 5 ply/3 layer plywood, use values for 4 ply. 
Effect of Green Lumber Framing on Diaphragms and Shear Walls: A recent study of wood 
structural panel shear walls (APA Report T2002-53) fabricated with wet lumber and tested when dry 
shows that shear stiffness is affected to a much larger degree than shear strength when compared to 
control specimens fabricated with dry lumber and tested when dry. The shear strength of walls 
fabricated with wet lumber showed negligible reductions (0-7 percent) when compared to control 
specimens. The shear stiffness of walls fabricated with wet lumber was always reduced when 
compared to control specimens. Observed reductions in stiffness were consistent with predicted 
stiffness reductions based on use of Eq. C12.2A and nail slip values specified in Table C12.2A. For 
example, measured deflection of a standard wall configuration at the shear wall factored unit shear 
value was approximately 2.5 times the deflection of the control specimen and predicted deflections 
were within 0.05 inches of the test deflection for both the fabricated wet specimen and control 
specimen. 
As a result of these tests, direct consideration of shear wall stiffness is recommended in lieu of 
applying shear wall strength reductions when wood structural panel shear walls are fabricated with 
wet lumber (e.g. moisture content > 19 percent). To address reduced shear stiffness for shear walls 
fabricated with wet lumber, story drift calculations should be based on en values for lumber with 
moisture content > 19 percent to determine compliance with allowable story drift limits of the 
Provisions. A similar relationship can be expected when analyzing the deflection of diaphragms. 
The designer should keep in mind that deflection equations are verified for walls with wood 
structural panel sheathing only and does not address the increased stiffness provided by finish 
materials such as gypsum and stucco. The CUREE-Caltech Woodframe project illustrated that 
finishes such as gypsum wallboard and stucco increase the stiffness of the walls. While these 
Wood Structure Design Requirements 
237 
deflection equations are currently the best estimate of wood structural panel wall deflection, actual 
wall deflections will likely be less than predicted deflections due to the presence of finish materials 
in typical wall construction. 
12.2.3.11 and 12.2.3.12. Tie-down devices should be based on cyclic tests of the connection to 
provide displacement capacity that allows rotation of the end post without significant reduction in 
the shear wall resistance. The tie-down device should be stronger than the lateral capacity of the 
wall so that the mechanism of failure is the sheathing fasteners and not a relatively brittle failure of 
the wall anchorage. For devices for which the published resistance is in allowable stress design 
values, the nominal strength shall be determined by multiplying the allowable design load by 1.3. 
The nominal strength of a tie-down device may be determined as the average maximum test load 
resisted without failing under cyclic loading. In that case, the average should be based on tests of at 
least three specimens. 
Calculations of deflection of shear walls should include the effects of crushing under the 
compression chord, uplift of the tension chord, slip in the tie-down anchor with respect to the post, 
and shrinkage effects of the platforms, which primarily consist of floor framing members. 
Movement associated with these variables can be significant and neglecting their contribution to the 
lateral displacement of the wall will results in a significant under-estimation of the deflection. 
Custom tie-down devices are permitted to be designed using methods for the particular materials 
used and AF&PA/ASCE 16 under alternative means and methods. 
Tie-down devices that permit significant vertical movement between the tie-down and the tie-down 
post can cause failure in the nails connecting the shear wall sheathing to the sill plate. High tension 
and tie-down rotation due to eccentricity can cause the bolts connecting the tie-down bracket to the 
tie-down post to pull through and split the tie-down post. Devices that permit such movement 
include heavily loaded, one-sided, bolted connections with small dimensions between elements 
resisting rotation due to eccentricity. Any device that uses over-drilled holes, such as most bolted 
connections, will also allow significant slip to occur between the device and the tie-down post 
before load is restrained. Both the NDS and the steel manual specify that bolt holes will be overdrilled 
as much as 1/16 in. (2 mm). This slip is what causes much of the damage to the nails 
connecting the sheathing to the sill plate. Friction between the tie-down post and the device cannot 
be counted on to resist load because relaxation in the wood will cause a loss of clamping and, 
therefore, a loss in friction over time. This is why all tests should be conducted with the bolts 
finger tight as opposed to tightening with a wrench. 
Cyclic tests of tie-down connections must follow a pattern similar to the sequential phased 
displacement (SPD) tests used by Dolan (1996) and Rose (1996). These tests used full wall 
assemblies and therefore induced deflection patterns similar to those expected during an earthquake. 
If full wall assembly tests are not used to test the tie-down devices, it must be shown that the 
expected rotation as well as tension and compression are used. This is to ensure that walls using the 
devices will be able to deform in the intended manner. This allows the registered design 
professional to consider compatibility of deformations when designing the structure. 
Splitting of the bottom plate of the shear walls has been observed in tests as well as in structures 
subjected to earthquakes. Splitting of plates remote from the end of the shear wall can be caused by 
the rotation of individual sheathing panels inducing upward forces in the nails at one end of the 
panel and downward forces at the other. With the upward forces on the nails and a significant 
distance perpendicular to the wall to the downward force produced by the anchor bolt, high crossgrain 
bending stresses occur. Splitting can be reduced or eliminated by use of large plate washers 
that are sufficiently stiff to reduce the eccentricity and by use of thicker sill plates. Thicker sill 
plates (3 in. nominal, 65 mm) are recommended for all shear walls for which Table 12.2-3a (or 12.2- 
3b) requires 3 in. nominal (65 mm) framing to prevent splitting due to close nail spacing. This is to 
help prevent failure of the sill plate due to high lateral loading and cross-grain bending. 
2 003 Commentary, Chapter 12 
238 
The tendency for the nut on a tie-down bracket anchor bolt to loosen significantly during cycled 
loading has been observed in some testing. One tested method of limiting the loosening is to apply 
adhesive between the nut and tie-down bolt. 
A logical load path for the structure must be provided so that the forces induced in the upper 
portions of the structure are transmitted adequately through the lower portions of the structure to the 
foundation. 
In the 2003 Provisions update cycle anchorage provisions were divided into two distinct subsections 
to separately address anchorage for uplift and anchorage for in-plane shear. The title section was 
clarified to address both traditional segmented shear walls and perforated shear walls. 
A prior Provisions requirement that nuts on both uplift anchors and in-plane shear anchors be 
prevented from loosening prior to covering the framing, was deleted. This provision was originally 
based on observed backing-off of nuts in a small number of cyclic tests of shear walls but in the 
large number of tests conducted since that time this phenomenon has not been observed to occur. It 
was felt that retaining the existing requirement for tightening the nuts prior to closing in the framing 
was sufficient to address this issue. 
A prior Provisions requirement for the nominal strength of a tiedown to be equal to or exceed the 
factored resistance of the shear wall times Oo 
/ 1.3, was replaced with simpler wording that has an 
equivalent effect and is intended primarily as a statement of design philosophy. The new language 
in Sec 12.2.3.11 only refers to the nominal strength of the tiedown and the nominal strength of the 
shear wall. Nominal strengths for typical nailed wood structural panel shear walls are set forth in 
Table 4.3A column B of AF&PA ASD/LRFD Supplement, Special Design Provisions for Wind and 
Seismic. In addition, similar language making the nominal strength of in-plane shear anchorage 
match the nominal strength values of the shear walls was added, to provide a basis for design of inplane 
shear connections that is consistent with requirements for uplift anchorage. The capacity-based 
nominal strength have been introduced primarily as a statement of design philosophy, with the intent 
of forcing sheathing nailing to be the controlling failure mechanism. The complexity of load paths in 
wood frame buildings suggest that additional study is needed to achieve reliable development of 
desired failure mechanisms. 
Plate washers are now specifically permitted to have a diagonal slot not exceeding 1-3/4 inches in 
length to facilitate placement within the width of the sill plate. 
12.2.3.14 Sheathing nails should be driven flush with the surface of the panel, and not further. This 
could result in the nail head creating a small depression in, but not fracturing, the first veneer. This 
requirement is imposed because of the significant reduction in capacity and ductility observed in 
shear walls constructed with over-driven nails. It is advised that the edge distance for sheathing 
nails be increased as much as possible along the bottom of the panel to reduce the potential for the 
nails to pull through the sheathing. 
12.3 GENERAL DESIGN REQUIREMENTS FOR ENGINEERED WOOD 
CONSTRUCTION 
Engineered construction for wood structures as defined by the Provisions encompasses all structures 
that cannot be classified as conventional construction. Therefore, any structure exceeding the height 
limitations or having braced walls spaced at intervals greater than those prescribed in Table 12.4-1 
or not conforming to the requirements in Sec. 12.4 must be engineered using standard design 
methods and principles of mechanics. Framing members in engineered wood construction are sized 
based on calculated capacities to resist the loads and forces imposed. Construction techniques that 
utilize wood for lateral force resistance in the form of diaphragms or shear walls are discussed 
further in Sec. 12.4. Limitations have been set on the use of wood diaphragms that are used in 
combination with concrete and masonry walls or where torsion is induced by the arrangement of the 
vertical resisting elements. A load path must be provided to transmit the lateral forces from the 
diaphragm through the vertical resisting elements to the foundation. It is important for the registered 
Wood Structure Design Requirements 
239 
design professional to follow the forces down, as for gravity loads, designing each connection and 
member along the load path. 
Although wood moment resisting frames are not specifically covered in the Provisions, they are not 
excluded by them. There are several technical references for their design, and they have been used 
in Canada, Europe, and New Zealand. Wood moment resisting frames are designed to resist both 
vertical loads and lateral forces. Detailing at columns to beam/girder connections is critical in 
developing frame action and must incorporate effects of member shrinkage. Detailed information 
can be obtained from the national wood research laboratories. There are many references that 
describe the engineering practices and procedures used to design wood structures that will perform 
adequately when subjected to lateral forces. The list at the end of this Commentary chapter gives 
some, but by no means all, of these. 
Deformation compatibility The registered design professional should visualize the deformed shape 
of the structure to ensure that the connections provide the necessary ductility to allow the probable 
deflection demand placed on the structure. Unlike steel or other metal structures, wood is not a 
ductile material and virtually all of the ductility achieved in the structure is in the connections. The 
planned failure mechanism of wood structures must be through the connections, including the 
nailing of structural panels; otherwise the failure will be brittle in nature. The philosophy of strong, 
elastic columns and yielding beams cannot be projected from steel to wood structures. To enable a 
wood structure to deform and dissipate energy during a seismic event, the connections must be the 
weak link in the structure and must be ductile. Recent earthquakes, such as that in Northridge, 
California, have shown failures due to the fact that consideration of deformation compatibility was 
neglected. 
As an example of a compatibility issue, consider the deformation compatibility between a tie-down 
connector to the tie-down post and the edge nailing of shear wall sheathing to the tie-down post and 
adjacent bottom plate. Recent testing and observations from the Northridge earthquake have 
suggested that the tie-down post experiences notable displacement before significant load can be 
carried through the tie-down connector. This is due, among other things, to the oversizing of the 
bolt holes in the tie-down post and the deformation and rotation of the tie-down bracket. Anchor 
bolts connecting the bottom plate to the foundation below tend to attempt to carry the shear wall 
uplift as the tie-down post moves. The sheathing, however, is nailed to both the bottom plate, which 
is held in place, and the tie-down post, which is being pulled up. The result is a large deformation 
demand being placed on the nails connecting the sheathing to the framing. This often results in the 
nails pulling out of the sheathing at the tie-down post corner and sometimes results in an unzipping 
effect where a significant portion of the remaining sheathing nailing fails as high loads cause one 
nailed connection to fail and move on to overstress the next nail. The most effective solution 
currently known is to limit the slip and deformation at the tie-down post by using a very stiff nailed 
or screwed tie-down. 
Because this is an area where understanding of compatibility issues is just starting to develop, the 
Sec. 12.3.2 provision uses the wording shall be considered in design in lieu of the originally 
proposed provision shall be made to ensure The intent is to provide guidance while not 
requiring the impossible. 
If necessary, the stiffness of the wood diaphragms and shear walls can be increased with the use of 
adhesives (if adhesives are to be used). However, it should be noted that there are no rational 
methods for determining deflections in diaphragms that are constructed with non-wood sheathing 
materials. If the nail stiffness values or shear stiffness of non-wood sheathing materials is 
determined in a scientific manner, such as through experimental cyclic testing, the calculations for 
determining the stiffness of shear panels will be considered validated. 

2 003 Commentary, Chapter 12 
240 
Limitation on forces contributed by concrete or masonry. Due to the significant difference in inplane 
stiffness between wood and masonry or concrete systems, the use of wood members to resist 
the seismic forces produced by masonry and concrete is not allowed. This is due to the probable 
torsional response such a structure will exhibit. There are two exceptions where wood can be 
considered to be part of the seismic-load-resisting system. The first is where the wood is in the form 
of a horizontal truss or diaphragm and the lateral loads do not produce rotation of the horizontal 
member. The second exception is in structures of two stories or less in height. In this case, the 
capacity of the wood shear walls will be sufficient to resist the lower magnitude loads imposed. 
Five restrictions are imposed on these structures to ensure that the structural performance will not 
include rotational response and that the drift will not cause failure of the masonry or concrete 
portions of the structure. 
Shearwalls and Diaphragms. Many wood-framed structures resist seismic forces by acting as a 
box system. The forces are transmitted through diaphragms, such as roofs and floors, to reactions 
provided by shear walls. The forces are, in turn, transmitted to the lower stories and to the final 
point of resistance, the foundations. A shear wall is a vertical diaphragm generally considered to act 
as a cantilever from the foundation. 
A diaphragm is a nearly horizontal structural unit that acts as a deep beam or girder when flexible in 
comparison to its supports and as a plate when rigid in comparison to its supports. The analogy to a 
girder is somewhat more appropriate since girders and diaphragms are made up as assemblies 
(American Plywood Association, 1991; Applied Technology Council, 1981). Sheathing acts as the 
web to resist the shear in diaphragms and is stiffened by the framing members, which also provide 
support for gravity loads. Flexure is resisted by the edge elements acting like flanges to resist 
induced tension or compression forces. The flanges may be top plates, ledgers, bond beams, or 
any other continuous element at the perimeter of the diaphragm. 
The flange (chord) can serve several functions at the same time, providing resistance to loads and 
forces from different sources. When it functions as the tension or compression flange of the 
girder, it is important that the connection to the web be designed to accomplish the shear 
transfer. Since most diaphragm flanges consist of many pieces, it is important that the splices be 
designed to transmit the tension or compression occurring at the location of the splice and to 
recognize that the direction of application of seismic forces can reverse. It should also be 
recognized that the shear walls parallel to the flanges may be acting with the flanges to distribute the 
diaphragm shears. When seismic forces are delivered at right angles to the direction considered 
previously, the flange becomes a part of the reaction system. It may function to transfer the 
diaphragm shear to the shear wall(s), either directly or as a drag strut between segments of shear 
walls that are not continuous along the length of the diaphragm. 
For shear walls, which may be considered to be deep vertical cantilever beams, the flanges are 
subjected to tension and compression while the webs resist the shear. It is important that the 
flange members, splices at intermediate floors, and the connection to the foundation be detailed 
and sized for the induced forces. 
The webs of diaphragms and shear walls often have openings. The transfer of forces around 
openings can be treated similarly to openings in the webs of steel girders. Members at the edges of 
openings have forces due to flexure and the higher web shear induced in them and the resultant 
forces must be transferred into the body of the diaphragm beyond the opening. 
In the past, wood sheathed diaphragms have been considered to be flexible by many registered 
design professionals and model code enforcement agencies. The newer versions of the model codes 
now recognize that the determination of rigidity or flexibility for determination of how forces will be 
distributed is dependent on the relative deformations of the horizontal and vertical force-resisting 
elements. Wood sheathed diaphragms in structures with wood frame shear walls with various types 
of sheathing may be relatively rigid compared with the vertical resisting system and, therefore, 
capable of transmitting torsional lateral forces. A diaphragm is considered to be flexible if its 
Wood Structure Design Requirements 
241 
deformation is two or more times that of the vertical force-resisting elements subjected to the same 
force. 
Discussions of these and other topics related to diaphragm and shear wall design, such as cyclic 
testing and pitched or notched diaphragms, may be found in the references. 
The capacity of shear walls must be determined either from tabulated values that are based on 
experimental results or from standard principles of mechanics. The tables of allowable values for 
shear walls sheathed with other than wood or wood-based structural-use panels were eliminated in 
the 1991 Provisions as a result of re-learning the lessons from past earthquakes and testing on the 
performance of structures sheathed with these materials during the Northridge earthquake. In the 
1997 Provisions values for capacity for shear walls sheathed with wood structural panels were 
reduced from monotonic test values by 10 percent to account for the reduction in capacity observed 
during cyclic tests. This decision was reviewed for the 2000 edition of the Provisions due to the 
availability of an expanded data set of test results. The reduction was removed for the 2000 
Provisions when the effect of the test loading protocol was determined to be the cause of the initial 
perceived reductions. Capacities for diaphragms were not reduced from the monotonic test values 
because the severe damage that occurred in shear walls has not been noted in diaphragms in recent 
earthquakes. 
The Provisions are based on assemblies having energy dissipation capacities which were recognized 
in setting the R factors. For diaphragms and shear walls utilizing wood framing, the energy 
dissipation is almost entirely due to nail bending. Fasteners other than nails and staples have not 
been extensively tested under cyclic load application. When screws or adhesives have been tested in 
assemblies subjected to cyclic loading, they have had a brittle mode of failure. For this reason, 
adhesives are prohibited for wood framed shear wall assemblies in SDC C and higher and only the 
tabulated values for nailed or stapled sheathing are recommended. If one wished to use shear wall 
sheathing attached with adhesives, as an alternate method of construction in accordance with Sec. 
1.1.2.5, caution should be used (Dolan and White, 1992; Foschi and Filiatrault, 1990). The 
increased stiffness will result in larger forces being attracted to the structure. The anchorage 
connections and adjoining assemblies must, therefore, be designed for these increased forces. Due 
to the brittle failure mode, these walls should be designed to remain elastic, similar to unreinforced 
masonry. The use of adhesives for attaching sheathing for diaphragms increases their stiffness, and 
could easily change the diaphragm response from flexible to rigid. 
Horizontal distribution of shear. The Provisions define when a diaphragm can be considered to 
be flexible or rigid. The purpose is to determine whether the diaphragm should have the loads 
proportioned according to tributary area or stiffness. For flexible diaphragms, the loads should be 
distributed according to tributary area whereas for rigid diaphragms, the loads should be distributed 
according to stiffness. 
The distribution of seismic forces to the vertical elements (shear walls) of the seismic-force-resisting 
system is dependent, first, on the stiffness of the vertical elements relative to that of the horizontal 
elements and, second, on the relative stiffness of the various vertical elements if they have varying 
deflection characteristics. The first issue is discussed in detail in the Provisions, which define when 
a diaphragm can be considered flexible or rigid and set limits on diaphragms that act in rotation or 
that cantilever. The second is largely an issue of engineering mechanics, but is discussed here 
because significant variations in engineering practice currently exist. 
In situations where a series of vertical elements of the seismic-force-resisting system are aligned in a 
row, seismic forces will distribute to the different elements according to their relative stiffness. 
Typical current design practice is to distribute seismic forces to a line of wood structural panel 
sheathed walls in proportion to the lengths of the wall segments such that each segment carries the 
same unit load. Wood structural panel sheathed wall segments without openings can generally be 
calculated to have a stiffness in proportion to the wall length when: the tie-down slip is ignored, the 
2 003 Commentary, Chapter 12 
242 
wood structural panel sheathing is selected from standard selection tables, and the aspect ratio limits 
of the Provisions are satisfied. For stiffness to be proportional to the wall length, the average load 
per nail for a given nail size must be approximately equal. Conversely, a wall could be stiffened by 
adding nails and reducing the calculated average load per nail. When including tie-down slip from 
anchors with negligible slip (1/16 in. [2 mm] or less), the assumption of wall stiffness proportional 
to length is still fairly reasonable. For larger tie-down slip values, wall stiffness will move towards 
being proportional to the square of the wall length; more importantly, however, the anchorage will 
start exhibiting displacement compatibility problems. For shear walls with aspect ratios higher than 
2/1, the stiffness is no longer in proportion to the length and equations are not available to 
reasonably calculate the stiffness. For a line of walls with variations in tie-down slip, chord framing, 
unit load per nail, or other aspects of construction, distribution of load to wall segments will need to 
be based on a deflection analysis. The shear wall and diaphragm deflection equations that are 
currently available are not always accurate. As testing results become available, the deflection 
calculation formulas will need to be updated and design assumptions for distribution of forces 
reviewed. 
Torsional diaphragm force distribution. A diaphragm is flexible when the maximum lateral 
deformation of the diaphragm is more than two times the average story drift. Conversely, a 
diaphragm will be considered rigid when the diaphragm deflection is equal to or less than two times 
the story drift. This is based on a model building code definition that applies to all materials. 
For flexible diaphragms, seismic forces should be distributed to the vertical force-resisting elements 
according to tributary area or simple beam analysis. Although rotation of the diaphragm may occur 
because lines of vertical elements have different stiffness, the diaphragm is not considered stiff 
enough to redistribute seismic forces through rotation. The diaphragm can be visualized as a singlespan 
beam supported on rigid supports. 
For diaphragms defined as rigid, rotational or torsional behavior is expected and results in 
redistribution of shear to the vertical force-resisting elements. Requirements for horizontal shear 
distribution are in Sec. 5.2.4. Torsional response of a structure due to irregular stiffness at any level 
within the structure can be a potential cause of failure. As a result, dimensional and diaphragm ratio 
limitations are provided for different categories of rotation. Also, additional requirements apply 
when the structure is deemed to have a torsional irregularity in accordance with Table 4.3-2, Item 1a 
or 1b. 
In order to understand limits placed on diaphragms acting in rotation, it is helpful to consider two 
different categories of diaphragms. Category I includes rigid diaphragms that rely on force transfer 
through rotation to maintain stability. An example would be an open front structure with shear walls 
on the other three sides. For this more structurally critical category, applicable limitations are: 
Diaphragm may not be used to resist forces contributed by masonry or concrete in structures over 
one story. 
The length of the diaphragm normal to the opening may not exceed 25 ft ( to perpendicular shear 
walls), and diaphragm L/b ratios are limited as noted. 
Additional limitations apply when rotation is significant enough to be considered a torsional 
irregularity. 
Category II includes rigid diaphragms that have two or more supporting shear walls in each of two 
perpendicular directions but, because the center of mass and center of rigidity do not coincide, 
redistribute forces to shear walls through rotation of the diaphragm. These can be further divided 
into Category IIA where the center of rigidity and mass are separated by a small portion of the 
structures least dimension and the magnitude of the rotation is on the order of the accidental 
rotation discussed in Sec. 5.2.4.2. For this level of rotation, an exception may result in no particular 
limitations being placed on diaphragm rotation for Category IIA. Category IIB, rigid diaphragms 
Wood Structure Design Requirements 
243 
with eccentricities larger than those discussed in Sec. 5.2.4.2, are subject to the following 
limitations: 
Diaphragm may not be used to resist forces contributed by masonry or concrete in structures over 
one story. 
Additional limitations apply when rotation is significant enough to be considered a torsional 
irregularity. 
Because flexible diaphragms have very little capacity for distributing torsional forces, further 
limitation of aspect ratios is used to limit diaphragm deformation such that rigid behavior will occur. 
The resulting deformation demand on the structure also is limited. Where diaphragm ratios are 
further limited, exceptions permit higher ratios where calculations demonstrate that higher 
diaphragm deflections can be tolerated. In this case, it is important to determine the effect of 
diaphragm rigidity on both the horizontal distribution and the ability of other structural elements to 
withstand resulting deformations. 
Proposals to prohibit wood diaphragms acting in rotation were advanced following the 1994 
Northridge earthquake. To date, however, the understanding is that the notable collapses in the 
Northridge earthquake occurred in part because of lack of deformation compatibility between the 
various vertical resisting elements rather than because of the inability of the diaphragm to act in 
rotation. 
Diaphragm cantilever. Limitations concerning diaphragms that cantilever horizontally past the 
outermost shear wall (or other vertical element) are related to but distinct from those imposed 
because of diaphragm rotation. Such diaphragms can be flexible or rigid and for rigid diaphragms 
can be Category I, IIA or IIB. Both the limitations based on diaphragm rotation (if applicable) and 
the following limit on diaphragm cantilever must be considered: 
Diaphragm cantilever may not exceed the lesser of 25 ft or two thirds of the diaphragm width. 
Relative stiffness of vertical elements. In situations where a series of vertical elements of the 
seismic-force-resisting system are aligned in a row, the forces will distribute to the different 
elements according to their relative stiffnesses. This behavior needs to be taken into account 
whether it involves a series of wood structural panel shear walls of different lengths, a mixture of 
wood structural panel shear walls with diagonal lumber or non-wood sheathed shear walls, or a 
mixture of wood shear walls with walls of some other material such as concrete or masonry. 
Diaphragm aspect ratio. The L/b for a diaphragm is intended to be the typical definition for aspect 
ratio. The diaphragm span, L, is measured perpendicular to the direction of applied force, either for 
the full dimension of the diaphragm or between supports as appropriate. The width, b, is parallel to 
the applied force (see Figure C12.3-1). 
Figure C12.3-1 Diaphragm dimension definitions. 
Single and double diagonally sheathed lumber diaphragms. Diagonally sheathed lumber 
diaphragms are addressed by the Provisions because they are still used for new construction in some 
regions. Shear resistance is based on a soft conversion from the model code allowable stress loads 
2 003 Commentary, Chapter 12 
244 
and capacities to Provisions strength loads for regions with high spectral accelerations. This will 
allow users in the western states, where this construction is currently being used, to continue with 
little or no change in requirements; at the same time, reasonable values are provided for regions with 
lower spectral 
Shear wall aspect ratio. The h/b for a shear wall is intended to be the typical definitions for aspect 
ratio. The h of the shear wall is the clear story height (see Figure C12.3-2). The alternate definition 
of aspect ratio is only to be used where specific design and detailing is provided for force transfer 
around the openings. It is required that the individual wall piers meet the aspect ratio requirement 
(see Figure C12.3-3) and that the overall perforated wall also meet the aspect ratio requirement. Use 
of the alternate definition involves the design and detailing of chord and collector elements around 
the opening, and often results in the addition of blocking, strapping, and special nailing. As noted, 
the design for force transfer around the opening must use a rational analysis and be in accordance 
with AF&PA/ASCE 16, which discusses design principles for shear walls, diaphragms, and 
boundary elements. 
In general, unit shear values for wood structural panel sheathing have been based on tests of shear 
wall panels with aspect ratios of 2/1 to 1/1. Narrower wall segments (that is, with aspect ratios 
greater than 2/1) have been a recent concern based on damage observations following the Northridge 
earthquake and based on results of recent research (Applied Technology Council, 1995; White and 
Dolan, 1996). In response, various limitations on aspect ratios have been proposed. In the 
Provisions, an aspect ratio adjustment, 2b/h, is provided to account for the reduced stiffness of 
narrow shear wall segments. This adjustment is based on a review of numerous tests of narrow 
aspect ratio walls by Technical Subcommittee 7. The maximum 3.5/1 aspect ratio is recommended 
based on constructability issues (placement of tie-downs) as well as reduced stiffness of narrower 
shear wall segments. 
Figure C12.3-2 Typical shear wall height-towidth 
ratio. 
Wood Structure Design Requirements 
245 
Figure C12.3-3 Alternate shear wall height-to-width 
ratio with design for force transfer around openings. 
Single and double diagonally sheathed lumber shear walls. Diagonally sheathed lumber shear 
walls are addressed by the Provisions because they are still used for new construction in some 
regions. Resistance values are based on a soft conversion from the model code allowable stress loads 
and capacities to Provisions strength loads for regions with high spectral accelerations. This will 
allow users in the western states, where this construction is currently being used, to continue with 
little or no change in requirements; at the same time, reasonable values are provided for regions with 
lower spectral accelerations. 
Perforated shear walls (PSW). In a traditional engineering approach for design of shear walls with 
openings, design force transfer around the openings involves developing a system of piers and 
coupling beams within the shear wall. Load paths for the shear and flexure developed in the piers 
and coupling beams generally require blocking and strapping extending from each corner of the 
opening to some distance beyond. This approach often results in shear wall detailing that is not 
practical to construct. 
The perforated shear wall approach utilizes empirically based reductions of wood structural panel 
shear wall capacities to account for the presence of openings that have not been specifically 
designed and detailed for moment resistance. This method accounts for the capacity that is inherent 
in standard construction, rather than relying on special construction requirements. It is not expected 
that sheathed wall areas above and below openings behave as coupling beams acting end to end, but 
rather that they provide local restraint at their ends. As a consequence significantly reduced 
capacities are attributed to interior perforated shear wall segments with limited overturning restraint. 
Example 1 and Example 2 provide guidance on the application of the perforated shear wall 
approach. 
2 003 Commentary, Chapter 12 
246 
Perforated Shear Wall Limitations. Perforated shear wall design provisions are applicable to 
wood structural panel shear walls having characteristics identified in this section. 
1. The requirement that perforated shear wall segments be provided at each end of the perforated 
shear wall ensures that a minimum length of full height sheathing, conforming to applicable 
aspect ratio limits, is included at each end of a perforated shear wall. 
2. A factored shear resistance not to exceed 0.64 klf, based on tabulated LRFD values, is provided 
to identify a point beyond which other means of shear wall design are likely to be more 
practical. Connection requirements associated with unadjusted shear resistance greater than 
0.64 klf will likely not be practical as other methods of shear wall design will be more efficient. 
3. Each perforated shear wall segment must satisfy the requirements for shear wall aspect ratios. 
The 2b/h adjustment for calculation of unadjusted factored shear resistance only applies when 
shear wall segments with h/b greater than 2:1 but not exceeding 3.5:1 are used in calculating 
perforated shear wall resistance. When shear wall segments with h/b greater than 2:1 are present 
in a perforated shear wall, but not utilized in calculation of perforated shear wall resistance, 
calculation of unadjusted factored shear resistance should not include the 2b/h adjustment. In 
many cases, due to the conservatism of the 2b/h adjustment, it is advantageous to simply ignore 
the presence of shear wall segments with h/b greater than 2:1 when calculating perforated shear 
wall resistance. 
4. No out-of-plane offsets are permitted in a perforated shear wall. While the limit on out-of-plane 
offsets is not unique to perforated shear walls, it is intended to clearly indicate that a perforated 
shear wall shall not have out-of-plane (horizontal) offsets. 
5. Collectors for shear transfer to each perforated shear wall segment provide for continuity 
between perforated shear wall segments. This is typically achieved through continuity of the 
wall double top plates or by attachment of perforated shear wall segments to a common load 
distributing element such as a floor or roof diaphragm. 
6. Uniform top-of-wall and bottom-of-wall elevations are required for use of the empirical shear 
adjustment factors. 
7. Limiting perforated shear wall height to 20 ft addresses practical considerations for use of the 
method as wall heights greater than 20 ft are uncommon. 
a. The width, L, of a perforated shear wall and widths L1, L2 and L3 of perforated shear 
wall segments are shown in Figure C12.3-4. In accordance with the limitations and 
anchorage requirements, perforated shear wall segments and overturning restraint 
must be provided at each end of the perforated shear wall. 
Perforated shear wall 
L1 L2 L3 
L 
h 
Overturning Restraint (each end) 
Figure C12.3-4 Perforated shear wall. 
Wood Structure Design Requirements 
247 
Perforated shear wall resistance. Opening adjustment factors are used to reduce shear wall 
resistance, based on the percent full-height sheathing and the maximum opening height ratio. 
Opening adjustment factors are based on the following empirical equation for shear capacity ratio, 
F, which relates the ratio of the shear capacity for a wall with openings to the shear capacity of a 
fully sheathed wall (Sugiyama, 1981): 
4 
3 2 
F 
r 
= 
- 
(C12.3-1a) 
1 
1 o 
i 
r 
h L 
A = 
+ S 
(C12.3-1b) 
where: 
r = sheathing area ratio, 
Ao = total area of openings, 
h = wall height, 
SLi = sum of the width of full-height sheathing. 
Agreement between Eq. C12.3-1a and tabulated opening adjustment factors is achieved by 
recognizing that the tabulated opening adjustment factors are: (1) derived based on an assumption 
that the height of all openings in a wall are equal to the maximum opening height; and, (2) applied to 
the sum of the widths of the shear wall segments meeting applicable height-to-width ratios. The 
assumption that the height of all openings in a wall are equal to the maximum opening height 
conservatively simplifies tabular presentation of shear capacity adjustment factors for walls with 
more than one opening height. 
Early verification of Eq. C12.3-1a was based on testing of one-third and full-scale shear wall 
assemblies (Yasumura, 1984; Sugiyama, 1994). More recently, substantial U.S. verification testing 
of the influence of openings on shear strength and stiffness has taken place (APA, 1996; Dolan and 
Johnson, 1996; Dolan and Heine, 1997; NAHB-RC, 1998) indicating shear wall performance is 
consistent with predictions of Eq. C12.3-1a. Results of cyclic testing indicate that the loss in 
strength due to cyclic loading is reduced for shear walls with openings, indicating good performance 
relative to that of shear walls without openings. Figure C12.3-5 provides a graphical summary of 
some recent U.S. verification testing. Data from monotonic tests of 12-ft shear walls (APA, 1996), 
monotonic and cyclic tests of long shear walls with unsymmetrically placed openings (Dolan and 
Johnson, 1996), and monotonic and cyclic tests of 16-ft and 20-ft shear walls with narrow wall 
segments (NAHB-RC, 1998). 
2 003 Commentary, Chapter 12 
248 
Sheathing Area Ratio, r 
0.0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 1.0 
Shear 
Capacity 
Ratio, F 
0.0 
0.1 
0.2 
0.3 
0.4 
0.5 
0.6 
0.7 
0.8 
0.9 
1.0 
F=r/(3-2r) 
APA - 1996 
Dolan - 1996 
NAHB-RC -1998 
Figure C12.3-5 Shear capacity ratios (actual and predicted). 
Eq. C12.3-1a for shear load ratio, F, has been shown to be a good approximation of the stiffness 
ratio of a wall with openings to that of a fully sheathed wall. Accordingly, the deflection of a 
perforated shear wall can be calculated as the deflection of an equivalent length fully sheathed wall, 
divided by the shear load ratio, F. 
The percent full-height sheathing and the maximum opening height ratio are used to determine an 
opening adjustment factor. Maximum opening height is the maximum vertical dimension of an 
opening within the perforated shear wall. A maximum opening height equal to the wall height is 
used where structural sheathing is not present above or below window openings or above door 
openings. The percent full-height sheathing is calculated as the sum of the widths of perforated 
shear wall segments divided by the total width of the shear wall. Sections sheathed full-height 
which do not meet aspect ratio limits for wood structural panel shear walls are not considered in 
calculation of percent full-height sheathing. 
PSW Anchorage and load path. Anchorage for uplift at perforated shear wall ends, shear, uplift 
between perforated shear wall ends, and compression chord forces are prescribed to address the nonuniform 
distribution of shear within a perforated shear wall. 
Prescribed forces for shear and uplift connections ensure that the capacity of the wall is governed by 
the sheathing to framing attachment (shear wall nailing) and not bottom plate attachment for shear 
and/or uplift. Shear and uplift forces approach the unadjusted factored shear resistance of the 
perforated shear wall segment as the shear load approaches the shear resistance of the perforated 
shear wall. A continuous load path to the foundation based on this requirement and consideration of 
other forces (for example, from the story above) shall be maintained. The magnitude of shear and 
uplift varies as a function of overturning restraint provided and aspect ratio of the shear wall 
segment. 

Wood Structure Design Requirements 
249 
Uplift anchorage at perforated shear wall ends. Anchorage for uplift forces due to overturning 
are required at each end of the perforated shear wall. A continuous load path to the foundation 
based on this requirement and consideration of other forces (for example, from the story above) shall 
be maintained. In addition, compression chords of perforated shear wall segments are required to 
transmit compression forces equal to the required tension chord uplift force. 
PSW Anchorage for in-plane shear. It is required that fastening be provided along the length of 
the sill plate of wall sections sheathed full-height to resist distributed shear, v, and uplift, t, forces. 
The resistance required for the shear connection is the average shear over the perforated shear wall 
segments, divided by the adjustment factor. This resistance will approach the unadjusted factored 
shear resistance of the wall as the shear wall demand approaches the maximum resistance. This 
shear fastening resistance conservatively accounts for the non-uniform distribution of shear within a 
perforated shear wall, since it represents the shear that can only be achieved when full overturning 
restraint is provided. 
Provisions require that distributed fastening for shear, v, and uplift, t, be provided over the length of 
full-height sheathed wall sections. With no other specific requirements, the fastening between the 
full height segments will be controlled by minimum construction fastening requirements. For 
bottom plates on wood platforms this would only require one 16-penny nail at 16 in. on center. In 
some cases, it may be preferable to extend a single bottom plate fastening schedule across the entire 
length of the perforated shear wall rather than to require multiple fastening schedules. 
Uplift anchorage between perforated shear wall ends. The resistance required for distributed 
uplift anchorage, t, is the same as the required shear resistance, v. The adequacy of the distributed 
uplift anchorage can be demonstrated using principles of mechanics and recent testing that 
determined the capacity of shear wall segments without uplift anchorage. A 4-ft wide shear wall 
segment with distributed anchorage of the base plate in lieu of an uplift anchor device provided 
about 25 percent of the resistance of a segment with uplift anchorage; an 8-ft wide shear wall 
segment resisted about 45 percent. When these are combined with the resistance adjustment factors, 
overturning resistance based on the unadjusted factored shear resistance is adequate for perforated 
shear wall segments with full height openings on each side. Conceptually the required distributed 
uplift resistance is intended to provide the same resistance that anchor bolts spaced at 2 ft on center 
provided for tested assemblies. While in the tested assemblies the bottom plates were fastened 
down, for design it is equally acceptable to fasten down the studs with a strap or similar device, 
since the studs will in turn restrain the bottom plate. 
PSW Load path. A continuous load path to the foundation is required for the uplift resistance, T; 
the compression resistance, C; the unit shear resistance, v; and the unit uplift resistance, t. 
Consideration of accumulated forces (for example, from the stories above) is required. Where shear 
walls occur at the same location at each floor (stack), accumulation of forces is reasonably 
straightforward. Where shear walls do not stack, attention will need to be paid to maintaining a load 
path for tie-downs at each end of the perforated shear wall, for compression resistance at each end of 
each perforated shear wall segment, and for distributed forces v and t at each perforated shear wall 
segment. Where ends of shear perforated shear wall segments occur over beams or headers, the 
beam or header will need to be checked for the vertical tension and compression forces in addition to 
gravity forces. Where adequate collectors are provided at lower floor shear walls, the total shear 
wall load need only consider the average shear in the perforated shear wall segments above, and not 
the average shear divided by the adjustment factor. 

2 003 Commentary, Chapter 12 
250 

Wood Structure Design Requirements 
251 

2 003 Commentary, Chapter 12 
252 
Wood Structure Design Requirements 
253 
Figure C12.3-6 Elevation for perforated shear wall Example 2.
2 003 Commentary, Chapter 12 
254 
Figure C12.3-7 Details for perforated shear wall Example 2.
Wood Structure Design Requirements 
255 
12.3.1 Framing. All framing that is designed as part of an engineered wood structure must be 
designed with connectors that are able to transfer the required forces between various components. 
These connectors can be either proprietary hardware or some of the more conventional connections 
used in wood construction. However, these connectors should be designed according to accepted 
engineering practice to ensure that they will have the capacity to resist the forces. The requirement of 
columns and posts being framed to full end bearing requires that the force transfer from the column to 
the base be accomplished through end grain bearing of the wood, not through placing the bolts or other 
connectors in shear. This requirement is included to ensure adequate capacity for transfer of the 
vertical forces due to both gravity and overturning moment. Alternatively, the connection can be 
designed to transfer the full loading through placing the bolts or other connectors in shear neglecting 
all possible bearing. 
The anchorage connections used in engineered wood construction must be capable of resisting the 
forces that will occur between adjacent members (beams and columns) and elements (diaphragms and 
shear walls). These connections can utilize proprietary hardware or be designed in accordance with 
principles of mechanics. Inadequate connections are often the cause of structural failures in wood 
structures, and the registered design professional is cautioned to use conservative values for allowable 
capacities since most published values are based on monotonic, not cyclic, load applications (U.S. 
Department of Agriculture, National Oceanic and Atmospheric Administration, 1971). Testing has 
shown that some one-sided bolted connections subject to cyclic loading, such as tie-down devices, do 
not perform well. This was substantiated by the poor performance of various wood frame elements in 
structures in the January 1994 Northridge earthquake. 
Concrete or masonry wall anchorages using toe nails or nails subject to withdrawal are prohibited by 
the Provisions. It has been shown that these types of connections are inadequate and do not perform 
well (U.S. Department of Agriculture, National Oceanic and Atmospheric Administration, 1971). 
Ledgers subjected to cross-grain bending or tension perpendicular to grain also have performed poorly 
in past earthquakes, and their use is now prohibited by the Provisions. 
12.4 CONVENTIONAL LIGHT-FRAME CONSTRUCTION 
The Provisions intend that a structure using conventional construction methods and complying with 
the requirements of this section be deemed capable of resisting the seismic forces imposed by the 
Provisions. 
Repetitive framing members such as joists, rafters, and studs together with sheathing and finishes 
comprise conventional light-frame construction. The subject of conventional construction is addressed 
in each of the model codes. It is acknowledged and accepted that, for the most part, the conventional 
construction provisions in the model codes concerning framing members and sheathing that carry 
gravity loads are adequate. This is due to the fact that the tables in the model codes giving allowable 
spans have been developed using basic principles of mechanics. For seismic lateral force resistance, 
however, experience has shown that additional requirements are needed. 
To provide lateral force resistance in vertical elements of structures, wall bracing requirements have 
been incorporated in conventional construction provisions of the model codes. With a few exceptions, 
these generally have been adequate for single family residences for which conventional construction 
requirements were originally developed. While the model building codes have been quite specific as 
to the type of bracing materials to be used and the amount of bracing required in any wall, no limits on 
the number or maximum separation between braced walls have been established. This section of the 
Provisions introduces the concept of mandating the maximum spacing of braced wall lines. By 
mandating the maximum spacing of braced wall lines and thereby limiting the lateral forces acting on 
these vertical elements, these revisions provide for a seismic-force-resisting system that will be less 
prone to overstressing and the requirements can be applied and enforced more uniformly than previous 
model building code requirements. While specific elements of light-frame construction may be 
calculated to be overstressed, there is typically a great deal of redundancy and uncounted resistance in 
such structures and they have generally performed well in past earthquakes. The experience in the 
2 003 Commentary, Chapter 12 
256 
Northridge earthquake was, however, less reassuring, especially for those residences relying on 
gypsum board or stucco for lateral force resistance. The light weight of conventional construction, 
together with the large energy dissipation capacity of the multiple fasteners used and inherent 
redundancy of the system are major factors in the observed good performance where wood or woodbased 
panels were used. 
12.4.1 Limitations 
12.4.1.1 General. The scope of this section specifically excludes prescriptive design of structures 
with concrete or masonry walls above the basement story, with the exception of veneer, in order to 
maintain the light weight of construction that the bracing requirements are based on. Wood braced 
wall panels and diaphragms as prescribed in this section are not intended to support lateral forces due 
to masonry or concrete construction. Prescriptive (empirical) design of masonry walls is allowed for 
in Chapter 11; however, design of structures combining masonry wall construction and wood roof and 
floor diaphragm construction must have an engineered design. In regions of high seismic activity, past 
earthquakes have demonstrated significant problems with structures combining masonry and wood 
construction. While engineered design requirements do address these problems, the prescriptive 
requirements in the model codes do not adequately address these problems. Masonry and concrete 
basement walls are permitted to be constructed in accordance with the requirements of the IRC. 
12.4.1.2 Irregular structures. This section was added to the 1997 Provisions to clarify the definition 
of irregular (unusually shaped) structures that would require the structure to be designed for the forces 
prescribed in Chapter 5 in accordance with the requirements of Sec. 12.3 and 12.4. The descriptions 
and diagrams provide the registered design professional with several typical irregularities that produce 
torsional response, or result in forces considered high enough to require an engineered design and 
apply only to structures assigned to Seismic Design Category C or D. 
Structures with geometric discontinuities in the lateral-force-resisting system have been observed to 
sustain more earthquake and wind damage than structures without discontinuities. They have also 
been observed to concentrate damage at the discontinuity location. For Seismic Design Categories C 
and D, this section translates applicable irregularities from Tables 4.3-2 and 4.3-3 into limitations on 
conventional light-frame construction. If the described irregularities apply to a given structure, it is 
required that either the entire structure or the non-conventional portions be engineered in accordance 
with the engineered design portions of the Provisions. The irregularities are based on similar model 
code requirements. While conceptually these are equally applicable to all seismic design categories, 
they are more readily accepted in areas of high seismic risk, where damage due to irregularities has 
been observed repeatedly. 
Application of engineered design to non-conventional portions rather than to the entire structure is a 
common practice in some regions. The registered design professional is left to judge the extent of the 
portion to be designed. This often involves design of the nonconforming element, force transfer into 
the element, and a load path from the element to the foundation. A nonconforming portion will 
sometimes have enough of an impact on the behavior of a structure to warrant that the entire seismicforce-
resisting system receives an engineered design. 
12.4.1.2.1 Out-of-plane offset. This limitation is based on Item 4 of Table 4.3-2 and applies when 
braced wall panels are offset out-of-plane from floor to floor. In-plane offsets are discussed in another 
item. Ideally braced wall panels would always stack above of each other from floor to floor with the 
length stepping down at upper floors as less length of bracing is required. 
Because cantilevers and set backs are very often incorporated into residential construction, the 
exception offers rules by which limited cantilevers and setbacks can be considered conventional. 
Floor joists are limited to 2 by 10 (actual: 12 by 93 in.; 38 by 235 mm) or larger and doubled at braced 
wall panel ends in order to accommodate the vertical overturning reactions at the end of braced wall 
panels. In addition the ends of cantilevers are attached to a common rim joist to allow for 
redistribution of load. For rim joists that cannot run the entire length of the cantilever, the metal tie is 
Wood Structure Design Requirements 
257 
intended to transfer vertical shear as well as to provide a nominal tension tie. Limitations are placed 
on gravity loads to be carried by cantilever or setback floor joists so that the joist strength will not be 
exceeded. The roof loads discussed are based on the use of solid sawn members where allowable 
spans limit the possible loads. Where engineered framing members such as trusses are used, gravity 
load capacity of the cantilevered or setback floor joists should be carefully evaluated. 
12.4.1.2.2 Unsupported diaphragm. This limitation is based in Item 1 of Table 4.3-2, and applies to 
open-front structures or portions of structures. The conventional construction bracing concept is based 
on using braced wall lines to divide a structure up into a series of boxes of limited dimension, with the 
seismic force to each box being limited by the size. The intent is that each box be supported by braced 
wall lines on all four sides, limiting the amount of torsion that can occur. The exception, which 
permits portions of roofs or floors to extend past the braced wall line, is intended to permit 
construction such as porch roofs and bay windows. Walls for which lateral resistance is neglected are 
allowed in areas where braced walls are not provided. 
12.4.1.2.3 Opening in wall below. This limitation is based on Item 4 of Table 4.3-3 and applies 
when braced wall panels are offset in-plane. Ends of braced wall panels supported on window or door 
headers can be calculated to transfer large vertical reactions to headers that may not be of adequate 
size to resist these reactions. The exception permits a 1 ft extension of the braced wall panel over a 4 
by 12 (actual: 32 by 113 in.; 89 by 286 mm) header on the basis that the vertical reaction is within a 45 
degree line of the header support and therefore will not result in critical shear or flexure. All other 
header conditions require an engineered design. Walls for which lateral resistance is neglected are 
allowed in areas where braced walls are not provided. 
12.4.1.2.4 Vertical offset in diaphragm. This limitation results from observation of damage that is 
somewhat unique to split-level wood frame construction. If floors on either side of an offset move in 
opposite directions due to earthquake or wind loading, the short bearing wall in the middle becomes 
unstable and vertical support for the upper joists can be lost, resulting in a collapse. If the vertical 
offset is limited to a dimension equal to or less than the joist depth, then a simple strap tie directly 
connecting joists on different levels can be provided, eliminating the irregularity. The IRC, Sec. 
502.6.1, provides requirements for tying of floor joists. 
12.4.1.2.5 Non-perpendicular walls. This limitation is based on Item 5 of Table 4.3-2 and applies to 
nonperpendicular braced wall lines. When braced wall lines are not perpendicular to each other, 
further evaluation is needed to determine force distributions and required bracing. 
12.4.1.2.6 Large diaphragm opening. This limitation is based on Item 3 of Table 4.3-2 and attempts 
to place a practical limit on openings in floors and roofs. Because stair openings are essential 
toresidential construction and have long been used without any report of life-safety hazards resulting, 
these are felt to be acceptable conventional construction. See Sec. 12.4.3.7 for detailing requirements 
for permitted openings. 
12.4.1.2.7 Stepped foundation. This limits a condition that can cause a torsional irregularity per 
Item 1 of Table 4.3-2. Where heights of braced wall panels vary significantly, distribution of lateral 
forces will also vary. If a structure on a hill is supported on 2-ft-high, braced cripple wall panels on 
one side and 8-ft-high panels on the other, torsion and redistribution of forces will occur. An 
engineered design for this situation is required in order to evaluate force distribution and provide 
adequate wall bracing and anchor bolting. This limitation applies specifically to walls from the 
foundation to the floor. While gable-end walls have similar variations in wall heights, this has not 
been observed to be a significant concern in conventional construction. See Sec. 12.4.3.6 for detailing 
requirements for permitted foundation stepping. 
12.4.2 Braced walls 
12.4.2.1 Spacing between braced wall lines. Table 12.4-1 prescribes the spacing of braced wall lines 
and number of stories permitted for conventional construction structures. Figures C12.4-1 and C12.4-
2 003 Commentary, Chapter 12 
258 
2 illustrate the basic components of the lateral bracing system. Information in Tables 12.4-1 and 12.4- 
2 was first included in the 1991 Provisions. 
12.4.2.2 Braced wall line sheathing. Table 12.4-2 prescribes the minimum length of bracing along 
each 25 ft (7.6 m) length of braced wall line. Total height of structures has been reduced to limit 
overturning of the braced walls so that significant uplift is not generally encountered. The height limit 
will accommodate 8 to 10 ft (2.4 to 3 m) story heights. 
12.4.3 Detailing requirements. The intent of this section is to rely on the traditional light-frame 
conventional construction materials and fastenings as prescribed in the references for this chapter. 
Braced wall panels are not required to be aligned vertically or horizontally (within the limits 
prescribed in Sec. 12.4.1) but stacking is desirable where possible. With the freedom provided for 
non-alignment it becomes important that a load path be provided to transfer lateral forces from upper 
levels through intermediate vertical and horizontal resisting elements to the foundation. Connections 
between horizontal and vertical resisting elements are prescribed. In structures two or three stories in 
height, it is desirable to have interior braced wall panels supported on a continuous foundation. See 
Figures C12.4-3 through C12.4-13 for examples of connections. 
The 1997 Provisions incorporated some of the wall anchorage, top plate, and braced wall panel 
connection requirements from the model building codes. These are included for completeness of the 
document and to clarify the requirement for the registered design professional. Additional 
requirements for foundations supporting braced wall panels has also been added to provide guidance 
and clarity for the registered design professional. 

Wood Structure Design Requirements 
259 
Figure C12.4-1 Acceptable one-story bracing example. 

2 003 Commentary, Chapter 12 
260 
Figure C12.4-2 Acceptable two-story bracing example. 

Wood Structure Design Requirements 
261 
Figure C12.4-3 Wall anchor detail. 
2 003 Commentary, Chapter 12 
262 
Figure C12.4-4 Double top splice. 
Figure C12.4-5 Single top splice. 

Wood Structure Design Requirements 
263 
Figure C12.4-6 Full bearing bottom plate. 
Figure C12.4-7 Exterior braced wall.
2 003 Commentary, Chapter 12 
264 
Figure C12.4-8 Interior braced wall at perpendicular joist. 
Figure C12.4-9 Interior braced wall at parallel joist. 

Wood Structure Design Requirements 
265 
Figure C12.4-10 Offset at interior braced wall. 

2 003 Commentary, Chapter 12 
266 
Figure C12.4-11 Diaphragm connection to braced wall below 
Figure C12.4-12 Post base detail. 
Wood Structure Design Requirements 
267 
Figure C12.4-13 Wood beam connection to post. 
12.4.3.4 Braced wall panel connections. The exception provided in this section of the Provisions is 
included due to the difficulty in providing a mechanism to transfer the diaphragm loads from a truss 
roof system to the braced wall panels of the top story. This problem has been considered by the 
Clackamas County, Oregon Building Codes Division, and an alternate to the CABO Building Code 
Sec. 402.10 was written in 1993, and revised September 5, 1995. The details shown in Figure C12.4- 
14 through C12.4-17 are provided as suggested methods for providing positive transfer of the lateral 
forces from the diaphragm through the web sections of the trusses to the top of the braced wall panels 
below. 
Figure C12.4-14 Suggested methods for transferring roof diaphragm load s to braced wall 
panels. 

2 003 Commentary, Chapter 12 
268 
Figure C12.4-15 Alternate gable end brace. 
Figure C12.4-16 Wall parallel to truss bracing detail. 

Wood Structure Design Requirements 
269 
Figure C12.4-17 Wall parallel to truss alternate bracing detail. 
REFERENCES 
American Institute of Timber Construction. 1994. Timber Construction Manual. New York, New 
York: John Wiley and Sons, Inc. 
American Forest & Paper Association. 1991. ANSI/NFoPA NDS-1991. National Design 
Specification for Wood Construction. Washington, D.C.: AF&PA. 
American Forest & Paper Association. 1996. Manual of Wood Construction: Load and Resistance 
Factor Design (LRFD). Washington, D.C.: AF&PA. 
American Plywood Association. 1991. Design/Construction GuideDiaphragms. Tacoma, 
Washington: APA. 
American Plywood Association. 1994. Northridge California Earthquake, T94-5. Tacoma, 
Washington.: APA. 
American Plywood Association. 1996. Proposed Cyclic Testing Standard for Shear Walls. Tacoma, 
Washington: APA. 
APA B The Engineered Wood Association. 1996. Wood structural panel shear walls with gypsum 
wallboard and window/door openings, Report 157. Tacoma, Washington: APA. 
ASCE Committee on Wood Load and Resistance Factor Design for Engineered Wood Construction. 
1988. A Pre Standard Report American Society of Civil Engineers. New York, New York: ASCE. 
ASCE. 1995. Standard For Load and Resistance Factor For Engineered Wood Construction. 
AF&PA/ASCE 16-95. American Society of Civil Engineers, New York, New York. 
Applied Technology Council. 1981. Guidelines for the Design of Horizontal Wood Diaphragms, 
ATC-7. Redwood City, California: ATC. 
2 003 Commentary, Chapter 12 
270 
Applied Technology Council. 1995. Cyclic Testing of Narrow Plywood Shear Walls, ATC-R-1. 
Redwood City, California: ATC. 
Breyer, Donald E. 1993. Design of Wood Structures, Third Edition. New York, New York: 
McGraw-Hill Book Company. 
Canadian Wood Council. 1990. Wood Design Manual. Ottawa: Canadian Wood Council. 
Canadian Wood Council. 1991. Wood Reference Handbook. Ottawa: Canadian Wood Council. 
Council of American Building Officials (CABO). 1995. One- and Two-Family Dwelling Code. 
Country Club Hills, Illinois: BOCA 
Department of the Army, Navy and Air Force. 1992. Seismic Design for Buildings, TM5-809-10 
(Tri-Services Manual). Washington, D.C.: U.S. Government Printing Office. 
Dolan, J.D., 1994. Proposed Test Method for Dynamic Properties of Connections Assembled with 
Mechanical Fasteners. ASTM Journal of Testing and Evaluation 22(6):542-547. 
Dolan, J. D., and M. W. White. 1992. Design Considerations for Using Adhesives in Shear Walls. 
ASCE Journal of Structural Engineering 118(12):3473-3480. 
Dolan, J. D. and Johnson, A. C. 1996. Monotonic Tests of Long Shear Walls with Openings 
Virginia Polytechnic Institute and State University, Timber Engineering Report No. TE-1996-001. 
Blacksburg, Virginia: VPISU. 
Dolan, J. D. and Johnson, A. C. 1996. Cyclic Tests of Long Shear Walls with Openings Virginia 
Polytechnic Institute and State University, Timber Engineering Report No. TE-1996-002. Blacksburg, 
Virginia: VPISU. 
Dolan, J. D., and C. P. Heine. 1997a. Monotonic tests of wood-frame shear walls with various 
openings and base restraint configurations, Timber Engineering Report TE-1997-001. Blacksburg: 
Virginia Polytechnic Institute and State University. 
Dolan, J. D., and C. P. Heine. 1997b. Sequential phased displacement cyclic tests of wood-frame 
shear walls with various openings and base restraint configurations, Timber Engineering Report TE- 
1997-002. Blacksburg: Virginia Polytechnic Institute and State University. 
Earthquake Engineering Research Institute. 1996. Northridge earthquake reconnaissance report, 
Chapter 6, Supplement C to Volume 11, pp. 125 et seq., Earthquake Spectra. 
Faherty, Keith F., and T. G. Williamson. 1989. Wood Engineering and Construction Handbook. 
New York, New York: McGraw-Hill. 
U.S. Department of Housing and Urban Development, 1984. HUD Minimum Property Standards, 
Vol. I, II, and III. Washington, D.C.: United States Government Printing Office. 
Forest Products Laboratory. 1986. Wood: Engineering Design Concepts. University Park: Materials 
Education Council, The Pennsylvania State University. 
Foschi, R. O., and A. Filiatrault. 1990. Performance Evaluation of 3M Scotch Grip Wood Adhesive 
5230 for the Static and Dynamic Design of Timber Shear Walls and Diaphragms, Vancouver: 
Department of Civil Engineering Report, University of British Columbia. 
Goetz, Karl-Heinz, Dieter Hoor, Karl Moehler, and Julius Natterer. 1989. Timber Design and 
Construction Source Book: A Comprehensive Guide to Methods and Practice. New York, New York: 
McGraw-Hill. 
Hoyle and Woeste. 1989. Wood Technology and Design of Structures. Iowa State University Press. 
International Conference of Building Officials (ICBO). 1994. Uniform Building Code. Whittier, 
California: ICBO. 
Wood Structure Design Requirements 
271 
Keenan, F. J. 1986. Limit States Design of Wood Structures. Morrison Hershfield Limited. 
NAHB Research Center. 1998. The performance of shear walls with narrow wall segments, reduced 
based restraint, and alternative framing methods. Upper Marlboro, Maryland. 
Rose, J. D., 1996. Research Report 158. Preliminary Testing of Wood Structural Panel Shear Walls 
Under Cyclic (Reversed) Loading. Tacoma, Washington: APA. 
Sherwood and Stroh. 1989. Wood-Frame House Construction in Agricultural Handbook 73. 
Washington, D.C.: U.S. Government Printing Office. 
Somayaji, Shan. 1992. Structural Wood Design. St. Paul, Minnesota: West Publishing Co. 
Stalnaker, Judith J., and E. C. Harris. 1996. Structural Design in Wood, Second Edition. New York, 
New York: McGraw-Hill. 
Sugiyama, H. 1981. The evaluation of shear strength of plywood-sheathed walls with openings, 
Wood Industry, 36-7, July. 
Sugiyama, H., and T. Matsumoto. 1994. Empirical equations for the estimation of racking strength 
of plywood-sheathed shear wall with openings, in Summaries of Technical Papers of Annual 
Meeting, Architectural Institute of Japan, September. 
U.S. Department of Agriculture, National Oceanic and Atmospheric Administration. 1971. San 
Fernando, California, Earthquake of February 9, 1971. Washington, D.C.: NOAA. 
White, M. W. and J. D. Dolan, 1996. Seismic Response of Timber Shear Walls. Part I: Aspect 
Ratios. Paper submitted for publication in ASCE Journal of Structural Engineering. 
Yasumura, M., and H. Sugiyama. 1984. Shear properties of plywood-sheathed wall panels with 
openings, in Transaction of the Architectural Institute of Japan, No. 338, April. 
Additional references are listed in John Peterson. 1983. Bibliography on Lumber and Wood Panel 
Diaphragms, Journal of Structural Engineering, Vol. 109 No. 12. American Society of Civil 
Engineers, New York, New York. 

2 003 Commentary, Chapter 12 
272 
This page is intentionally left blank. 

 273 
Chapter 13 Commentary 
SEISMICALLY ISOLATED STRUCTURE DESIGN REQUIREMENTS 
13.1 GENERAL 
Seismic isolation, commonly referred to as base isolation, is a design concept based on the premise that 
a structure can be substantially decoupled from potentially damaging earthquake motions. By 
substantially decoupling the structure from the ground motion, the level of response in the structure can 
be reduced significantly from the level that would otherwise occur in a conventional, fixed-base 
building. . 
The potential advantages of seismic isolation and the recent advancements in isolation-system products 
already have led to the design and construction of over 200 seismically isolated buildings and bridges in 
the United States. A significant amount of research, development, and application activity has occurred 
over the past 20 years. The following references provide a summary of some of the work that has been 
performed: Applied Technology Council (ATC, 1986, 1993, and 2002), ASCE Structures Congress 
(ASCE, 1989, 1991, 1993, and 1995), EERI Spectra (EERI, 1990), Skinner, et al. (1993), U.S. 
Conference on Earthquake Engineering (1990 and 1994), and World Conference on Earthquake 
Engineering (1988, 1992, 1996, and 2000). 
In the mid-1980s, the initial applications identified a need to supplement existing codes with design 
requirements developed specifically for seismically isolated buildings. Code development work 
occurred throughout the late 1980s. The status of U.S. seismic isolation design requirements as of May 
2003 is as follows: 
1. In late 1989, the Structural Engineers Association of California (SEAOC) State Seismology 
Committee adopted an Appendix to Chapter 2 of the SEAOC Blue Book entitled, General 
Requirements for the Design and Construction of Seismic-Isolated Structures. These requirements 
were submitted to the International Conference of Building Officials (ICBO) and were adopted by 
ICBO as an appendix of the 1991 Uniform Building Code (UBC). The most current version of these 
regulations may be found in the ASCE-7-02 (ASCE, 2003) and the 2003 International Building 
Code (ICC, 2003).. 
2. In 1991 the Federal Emergency Management Agency (FEMA) initiated a 6-year program to develop 
a set of nationally applicable guidelines for seismic rehabilitation of existing buildings. These 
guidelines (known as the NEHRP Guidelines for the Seismic Rehabilitation of Buildings) were 
published as FEMA 273. In 2000, FEMA 273 was republished, with minor amendments, as FEMA 
356, Prestandard and Commentary for the Seismic Rehabilitation of Buildings. The design and 
analysis methods of the NEHRP Guidelines and the FEMA Prestandard parallel closely methods 
required by the NEHRP Recommended Provisions for new buildings, except that more liberal 
design is permitted for the superstructure of a rehabilitated building. 
A general concern has long existed regarding the applicability of different types of isolation systems. 
Rather than addressing a specific method of base isolation, the Provisions provides general design 
requirements applicable to a wide range of possible seismic isolation systems. 
Although remaining general, the design requirements rely on mandatory testing of isolation-system 
hardware to confirm the engineering parameters used in the design and to verify the overall adequacy of 
the isolation system. Some systems may not be capable of demonstrating acceptability by test and, 
consequently, would not be permitted. In general, acceptable systems will: (1) remain stable for
2003 Commentary, Chapter 13 
274 
required design displacements, (2) provide increasing resistance with increasing displacement, (3) not 
degrade under repeated cyclic load, and (4) have quantifiable engineering parameters (such as forcedeflection 
characteristics and damping). 
Conceptually, there are four basic types of isolation system force-deflection relationships. These 
idealized relationships are shown in Figure C13.1-1 with each idealized curve having the same design 
displacement, DD, for the design earthquake. A linear isolation system is represented by Curve A and 
has the same isolated period for all earthquake load levels. In addition, the force generated in the 
superstructure is directly proportional to the displacement across the isolation system. 
A hardening isolation system is represented by Curve B. This system is soft initially (long effective 
period) and then stiffens (effective period shortens) as the earthquake load level increases. When the 
earthquake load level induces displacements in excess of the design displacement in a hardening system, 
the superstructure is subjected to higher forces and the isolation system to lower displacements than a 
comparable linear system. 
A softening isolation system is represented by Curve C. This system is stiff initially (short effective 
period) and softens (effective period lengthens) as the earthquake load level increases. When the 
earthquake load level induces displacements in excess of the design displacement in a softening system, 
the superstructure is subjected to lower forces and the isolation system to higher displacements than a 
comparable linear system. 
A sliding isolation system is represented by Curve D. This system is governed by the friction force of 
the isolation system. Like the softening system, the effective period lengthens as the earthquake load 
level increases and loads on the superstructure remain constant. 
The total system displacement for extreme displacement of the sliding isolation system, after repeated 
earthquake cycles, is highly dependent on the vibratory characteristics of the ground motion and may 
exceed the design displacement, DD. Consequently, minimum design requirements do not adequately 
define peak seismic displacement for seismic isolation systems governed solely by friction forces. 
13.1.1 Scope. The requirements of Chapter 13 provide isolator design displacements, shear forces for 
structural design, and other specific requirements for seismically isolated structures. All other design 
requirements including loads (other than seismic), load combinations, allowable forces and stresses, and 
horizontal shear distribution are covered by the applicable sections of the Provisions for conventional, 
fixed-base structures. 
Figure C13.1-1 Idealized force-deflection relationships for 
isolation systems (stiffness effects of sacrificial windrestraint 
systems not shown for clarity). 
Seismically Isolated Structure Design Requirements 
275 
13.2 GENERAL DESIGN REQUIREMENTS 
13.2.1 Occupancy importance factor. Ideally, most of the lateral displacement of an isolated 
structure will be accommodated by deformation of the isolation system rather than distortion of the 
structure above. Accordingly, the lateral-load-resisting system of the structure above the isolation 
system should be designed to have sufficient stiffness and strength to avoid large, inelastic 
displacements. For this reason, the Provisions contains criteria that limit the inelastic response of the 
structure above the isolation system. Although damage control for the design-level earthquake is not an 
explicit objective of the Provisions, an isolated structure designed to limit inelastic response of the 
structural system also will reduce the level of damage that would otherwise occur during an earthquake. 
In general, isolated structures designed in conformance with the Provisions should be able: 
1. To resist minor and moderate levels of earthquake ground motion without damage to structural 
elements, nonstructural components, or building contents; and 
2. To resist major levels of earthquake ground motion without failure of the isolation system, without 
significant damage to structural elements, without extensive damage to nonstructural components, 
and without major disruption to facility function. 
The above performance objectives for isolated structures considerably exceed the performance 
anticipated for fixed-base structures during moderate and major earthquakes. Table C13.2-1 provides a 
tabular comparison of the performance expected for isolated and fixed-base structures designed in 
accordance with the Provisions. Loss of function is not included in Table C13.2-1. For certain (fixedbase) 
facilities, loss of function would not be expected to occur until there is significant structural 
damage causing closure or restricted access to the building. In other cases, the facility could have only 
limited or no structural damage but would not be functional as a result of damage to vital nonstructural 
components and contents. Isolation would be expected to mitigate structural and nonstructural damage 
and to protect the facility against loss of function. 
Table C13.2-1 Protection Provided by NEHRP Recommended Provisions for Minor, 
Moderate, and Major Levels of Earthquake Ground Motion 
Earthquake Ground Motion Level 
Risk Category 
Minor 
Moderate 
Major 
Life safetya 
F, I 
F, I 
F, I 
Structural damageb 
F, I 
F, I 
I 
Nonstructural damagec (contents damage) 
F, I 
I 
I 
a Loss of life or serious injury is not expected. 
b Significant structural damage is not expected. 
c Significant nonstructural (contents) damage is not expected. 
F indicates fixed base; I indicates isolated. 
13.2.3.1 Design spectra. Site-specific design spectra must be developed for both the design earthquake 
and the maximum considered earthquake if the structure is located at a site with S1 greater than 0.60 or 
on a Class F site. All requirements for spectra are in Sec. 3.3 and 3.4. 
13.2.4 Procedure selection. The design requirements permit the use of one of three different analysis 
procedures for determining the design-level seismic loads. The first procedure uses a simple, lateralforce 
formula (similar to the lateral-force coefficient now used in conventional building design) to 
prescribe peak lateral displacement and design force as a function of spectral acceleration and isolatedbuilding 
period and damping. The second and third methods, which are required for geometrically 
complex or especially flexible buildings, rely on dynamic analysis procedures (either response spectrum 
or time history) to determine peak response of the isolated building. 
2003 Commentary, Chapter 13 
276 
The three procedures are based on the same level of seismic input and require a similar level of 
performance from the building. There are benefits in performing a more complex analysis in that 
slightly lower design forces and displacements are permitted as the level of analysis becomes more 
sophisticated. The design requirements for the structural system are based on the design earthquake, a 
severe level of earthquake ground motion defined as two-thirds of the maximum considered earthquake. 
The isolation systemincluding all connections, supporting structural elements, and the gapis 
required to be designed (and tested) for 100 percent of maximum considered earthquake demand. 
Structural elements above the isolation system are not required to be designed for the full effects of the 
design earthquake , but may be designed for slightly reduced loads (that is, loads reduced by a factor of 
up to 2.0) if the structural system has sufficient ductility, etc., to respond inelastically without sustaining 
significant damage. A similar fixed-base structure would be designed for loads reduced by a factor of 8 
rather than 2. 
This section delineates the requirements for the use of the equivalent lateral force procedure and 
dynamic methods of analysis. The limitations on the simplified lateral-force design procedure are quite 
severe at this time. Limitations cover the site location with respect to active faults; soil conditions of the 
site, the height, regularity and stiffness characteristics of the building; and selected characteristics of the 
isolation system. Response-history analysis is required to determine the design displacement of the 
isolation system (and the structure above) for the following isolated structures: 
1. Isolated structures with a nonlinear isolation system including, but not limited to, isolation 
systems utilizing friction or sliding surfaces, isolation systems with effective damping values greater 
than about 30 percent of critical, isolation systems not capable of producing a significant restoring 
force, and isolation systems that restrain or limit extreme earthquake displacement; 
2. Isolated structures with a nonlinear structure (above the isolation system) including, but not 
limited to, structures designed for forces that are less than those specified by the Provisions for 
essentially-elastic design; and 
3. Isolated structures located on Class F site (that is, very soft soil). 
Lower-bound limits on isolation system design displacements and structural-design forces are specified 
by the Provisions in Sec. 13.4 as a percentage of the values prescribed by the equivalent-lateral-force 
design formulas, even when dynamic analysis is used as the basis for design. These lower-bound limits 
on key design parameters ensure consistency in the design of isolated structures and serve as a safety 
net against gross under-design. Table C13.2-2 provides a summary of the lower-bound limits on 
dynamic analysis specified by the Provisions. 
13.2.4.3 Variations in material properties: For analysis, the mechanical properties of seismic 
isolators are generally based on values provided by isolator manufacturers. The properties are evaluated 
by prototype testing, which often occurs shortly after the isolators have been manufactured, and checked 
with respect to the values assumed for design. Unlike conventional materials whose properties do not 
vary substantially with time, seismic isolators are composed of materials whose properties will generally 
vary with time. Because (a) mechanical properties can vary over the life span of a building, and (b) the 
testing protocol of Section 13.6 cannot account for the effects of aging, contamination, scragging 
(temporary degradation of mechanical properties with repeated cycling), temperature, velocity effects, 
and wear, the engineer-of-record must account for these effects by explicit analysis. One strategy for 
accommodating these effects makes use of property modification factors, which was introduced by 
Constantinou et al. (1999) in the AASHTO Guide Specification for Seismic Isolation Design 
(AASHTO, 1999). Constantinou et al. (1999) also provides information on variations in material 
properties for sliding isolation systems. Thompson et al. (2000) and Morgan et al. (2001) provide 
information on variations in material properties for elastomeric bearings. 

Seismically Isolated Structure Design Requirements 
277 
Table C13.2-2 Lower-Bound Limits on Dynamic Procedures Specified in Relation to 
ELF Procedure Requirements 
Dynamic Procedure 
Design Parameter 
ELF Procedure 
Response 
Spectrum 
Response 
History 
Design displacement  DD 
DD = (g/4p2)(SD1TD/BD) 
B 
B 
Total design displacement - DT 
DT = 1.1D 
= 0.9DT 
= 0.9DT 
Maximum displacement  DM 
DM = (g/4p2)(SM1TM/BM) 
B 
B 
Total maximum displacement - DTM 
DTM = 1.1DM 
= 0.8DTM 
= 0.8DTM 
Design shear  Vb 
(at or below the isolation system) 
Vb = kDmaxDD 
= 0.9Vb 
= 0.9Vb 
Design shear  Vs 
(regular superstructure) 
Vs = kDmaxDD/RI 
= 0.8Vs 
= 0.6Vs 
Design shear  Vs 
(irregular superstructure) 
Vs = kDmaxDDRI 
= 1.0Vs 
= 0.8Vs 
Drift (calculated using RI for Cd) 
0.015hsx 
0.015hsx 
0.020hsx 
13.2.5 Isolation system 
13.2.5.1 Environmental conditions. Environmental conditions that may adversely affect isolation 
system performance should be thoroughly investigated. Significant research has been conducted on the 
effects of temperature, aging, etc., on isolation systems since the 1970s in Europe, New Zealand, and the 
United States. 
13.2.5.2 Wind forces. Lateral displacement over the depth of the isolator zone resulting from wind 
loads should be limited to a value similar to that required for other story heights. 
13.2.5.3 Fire resistance. In the event of a fire, the isolation system should be capable of supporting the 
weight of the building, as required for other vertical-load-supporting elements of the structure, but may 
have diminished functionality for lateral (earthquake) load. 
13.2.5.4 Lateral-restoring force. The isolation system should be configured with a lateral-restoring 
force sufficient to avoid significant residual displacement as a result of an earthquake, such that the 
isolated structure will not have a stability problem so as to be in a condition to survive aftershocks and 
future earthquakes. 
13.2.5.5 Displacement restraint. The use of a displacement restraint is not encouraged by the 
Provisions. Should a displacement restraint system be implemented, explicit analysis of the isolated 
structure for maximum considered earthquake is required to account for the effects of engaging the 
displacement restraint. 
13.2.5.6 Vertical-load stability. The vertical loads to be used in checking the stability of any given 
isolator should be calculated using bounding values of dead load and live load and the peak earthquake 
demand of the maximum considered earthquake. Since earthquake loads are reversible in nature, peak 
earthquake load should be combined with bounding values of dead and live load in a manner which 
produces both the maximum downward force and the maximum upward force on any isolator. Stability 
of each isolator should be verified for these two extreme values of vertical load at peak maximum 
considered earthquake displacement of the isolation system. 
2003 Commentary, Chapter 13 
278 
13.2.5.7 Overturning. The intent of this requirement is to prevent both global structural overturning 
and overstress of elements due to local uplift. Uplift in a braced frame or shear wall is acceptable so 
long as the isolation system does not disengage from its horizontal-resisting connection detail. The 
connection details used in some isolation systems are such that tension is not permitted on the system. 
If the tension capacity of an isolation system is to be utilized to resist uplift forces, then component tests 
should be performed to demonstrate the adequacy of the system to resist tension forces at the design 
displacement. 
13.2.5.8 Inspection and replacement. Although most isolation systems will not need to be replaced 
after an earthquake, it is good practice to provide for inspection and replacement. After an earthquake, 
the building should be inspected and any damaged elements should be replaced or repaired. It is 
advised that periodic inspections be made of the isolation system. 
13.2.5.9 Quality control. A test and inspection program is necessary for both fabrication and 
installation of the isolation system. Because base isolation is a developing technology, it may be 
difficult to reference standards for testing and inspection. Reference can be made to standards for some 
materials such as elastomeric bearings (ASTM D 4014). Similar standards are required for other 
isolation systems. Special inspection procedures and load testing to verify manufacturing quality should 
be developed for each project. The requirements will vary with the type of isolation system used. 
13.2.6 Structural system 
13.2.6.1 Horizontal distribution of force 
13.2.6.2 Building separations. A minimum separation between the isolated structure and a rigid 
obstruction is required to allow free movement of the superstructure in all lateral directions during an 
earthquake. Provision should be made for lateral motion greater than the design displacement, since the 
exact upper limit of displacement cannot be precisely determined. 
13.2.7 Elements of structures and nonstructural components. To accommodate the differential 
movement between the isolated building and the ground, provision for flexible utility connections 
should be made. In addition, rigid structures crossing the interface (such as stairs, elevator shafts and 
walls) should have details to accommodate differential motion at the isolator level without sustaining 
damage sufficient to threaten life safety. 
13.3 EQUIVALENT LATERAL FORCE PROCEDURE 
13.3.2 Minimum lateral displacements. The lateral displacement given by Eq. 13.3-1 approximates 
peak design earthquake displacement of a single-degree-of-freedom, linear-elastic system of period, TD, 
and equivalent viscous damping, D, and the lateral displacement given by Eq. 13.3-3 approximates 
peak maximum considered earthquake displacement of a single-degree-of-freedom, linear-elastic system 
of period, TM, and equivalent viscous damping, DM. 
Equation 13.3-1 is an estimate of peak displacement in the isolation system for the design earthquake. 
In this equation, the spectral acceleration term, SD1, is the same as that required for design of a 
conventional fixed-base structure of period, TD. A damping term, BD, is used to decrease (or increase) 
the computed displacement when the equivalent damping coefficient of the isolation system is greater 
(or smaller) than 5 percent of critical damping. Values of coefficient BD (or BM for the maximum 
considered earthquake) are given in Table 13.3-1 for different values of isolation system damping, D 
(or M). 
A comparison of values obtained from Eq. 13.3-1 and those obtained from nonlinear time-history 
analyses are given in Kircher et al. (1988) and Constantinou et al. (1993). 
Consideration should be given to possible differences in the properties of the isolation system used for 
design and the properties of isolation system actually installed in the building. Similarly, consideration 
should be given to possible changes in isolation system properties due to different design conditions or 
load combinations. If the true deformational characteristics of the isolation system are not stable or vary 
Seismically Isolated Structure Design Requirements 
279 
with the nature of the load (being rate-, amplitude-, or time-dependent), the design displacements should 
be based on deformational characteristics of the isolation system that give the largest possible deflection 
(kDmin), the design forces should be based on deformational characteristics of the isolation system that 
give the largest possible force (kDmax), and the damping level used to determine design displacements 
and forces should be based on deformational characteristics of the isolation system that represent the 
minimum amount of energy dissipated during cyclic response at the design level. 
The configuration of the isolation system for a seismically isolated building or structure should be 
selected in such a way as to minimize any eccentricity between the center of mass of the superstructure 
and the center of rigidity of the isolation system. In this way, the effect of torsion on the displacement 
of isolation elements will be reduced. As for conventional structures, allowance for accidental 
eccentricity in both horizontal directions must be considered. Figure C13.3-1 defines the terminology 
used in the Provisions. Equation 13.3-5 (or Eq. 13.3-6 for the maximum considered earthquake) 
provides a simplified formulae for estimating the response due to torsion in lieu of a more refined 
analysis. The additional component of displacement due to torsion increases the design displacement at 
the corner of the structure by about 15 percent (for a perfectly square building in plan) to about 30 
percent (for a very long, rectangular building) if the eccentricity is 5 percent of the maximum plan 
dimension. Such additional displacement, due to torsion, is appropriate for buildings with an isolation 
system whose stiffness is uniformly distributed in plan. Isolation systems that have stiffness 
concentrated toward the perimeter of the building or certain sliding systems that minimize the effects of 
mass eccentricity will have reduced displacements due to torsion. The Provisions permits values of DT 
as small as 1.1DD, with proper justification. 
Figure C13.3-1 Displacement terminology. 
2003 Commentary, Chapter 13 
280 
13.3.3 Minimum lateral forces. Figure C13.3-2 defines the terminology below and above the 
isolation system. Equation 13.3-7 gives peak seismic shear on all structural components at or below the 
seismic interface without reduction for ductile response. Equation 13.3-8 specifies the peak seismic 
shear for design of structural systems above the seismic interface. For structures that have appreciable 
inelastic-deformation capability, this equation includes an effective reduction factor of up to 2 for 
response beyond the strength-design level. 
The basis for the reduction factor is that the design of the structural system is based on strength-design 
procedures. A factor of at least 2 is assumed to exist between the design-force level and the true-yield 
level of the structural system. An investigation of 10 specific buildings indicated that this factor varied 
between 2 and 5 (ATC, 1982). Thus, a reduction factor of 2 is appropriate to ensure that the structural 
system remains essentially elastic for the design earthquake . 
In Sec. 13.3.3.2, the limitations given on VS ensure that there is at least a factor of 1.5 between the 
nominal yield level of the superstructure and (1) the yield level of the isolation system, (2) the ultimate 
capacity of a sacrificial wind-restraint system which is intended to fail and release the superstructure 
during significant lateral load, or (3) the break-away friction level of a sliding system. 
These limitations are essential to ensure that the superstructure will not yield prematurely before the 
isolation system has been activated and significantly displaced. 
The design shear force, VS, specified by the requirements of this section ensures that the structural 
system of an isolated building will be subjected to significantly lower inelastic demands than a 
conventionally designed structure. Further reduction in VS, such that the inelastic demand on a 
seismically isolated structure would be the same as the inelastic demand on a conventionally designed 
structure, was not considered during development of these requirements but may be considered in the 
future. 
If the level of performance of the isolated structure is desired to be greater than that implicit in these 
requirements, then the denominator of Eq. 13.3-8 may be reduced. Decreasing the denominator of Eq. 
13.3-8 will lessen or eliminate inelastic response of the superstructure for the design-basis event. 
13.3.4 Vertical distribution of forces. Equation 13.3-9 describes the vertical distribution of lateral 
force based on an assumed triangular distribution of seismic acceleration over the height of the structure 
Figure C13.3-2 Isolation system 
terminology
Seismically Isolated Structure Design Requirements 
281 
above the isolation interface. Constantinou et al. (1993) provides a good summary of recent work which 
demonstrates that this vertical distribution of force will always provide a conservative estimate of the 
distributions obtained from more detailed, nonlinear analysis studies. 
13.3.5 Drift limits. The maximum story drift permitted for design of isolated structures varies 
depending on the method of analysis used, as summarized in Table C13.3-1. For comparison, the drift 
limits prescribed by the Provisions for fixed-base structures also are summarized in Table C13.3-1. 
Table C13.3-1 Comparison of Drift Limits for Fixed-Base and Isolated Structures 
Structure 
Seismic Use Group 
Fixed-Base 
Isolated 
I 
0.025hsx/(Cd/R) 
0.015hsx 
II 
0.020hsx/(Cd/R) 
0.015hsx 
Buildings (other than 
masonry) four stories or 
less in height with 
component drift design 
III 
0.015hsx/(Cd/R) 
0.015hsx 
I 
0.020hsx/(Cd/R) 
0.015hsx 
II 
0.015hsx/(Cd/R) 
0.015hsx 
Other (non-masonry) 
buildings 
III 
0.010hsx/(Cd/R) 
0.015hsx 
Drift limits in Table C13.3-1 are divided by Cd/R for fixed-base structures since displacements 
calculated for lateral loads reduced by R are factored by Cd before checking drift. The Cd term is used 
throughout the Provisions for fixed-base structures to approximate the ratio of actual earthquake 
response to response calculated for reduced forces. Generally, Cd is 1/2 to 4/5 the value of R. For 
isolated structures, the RI factor is used both to reduce lateral loads and to increase displacements 
(calculated for reduced lateral loads) before checking drift. Equivalency would be obtained if the drift 
limits for both fixed-base and isolated structures were based on their respective R factors. It may be 
noted that the drift limits for isolated structures are generally more conservative than those for 
conventional, fixed-base structures, even when fixed-base structures are designed as Seismic Use Group 
III buildings. 
13.4 DYNAMIC PROCEDURES 
This section specifies the requirements and limits for dynamic procedures. The design displacement 
and force limits on response spectrum and response history procedures are given in Table C13.2-1. 
A more-detailed or refined study can be performed in accordance with the analysis procedures described 
in this section. The intent of this section is to provide procedures which are compatible with the 
minimum requirements of Sec. 13.3. Reasons for performing a more refined study include: 
1. The importance of the building. 
2. The need to analyze possible structure/isolation-system interaction when the fixed-base period of 
the building is greater than one third of the isolated period. 
3. The need to explicitly model the deformational characteristics of the lateral-force-resisting system 
when the structure above the isolation system is irregular. 
4. The desirability of using site-specific ground-motion data, especially for soft soil types (Site Class 
F) or for structures located where S1 is greater than 0.60. 
5. The desirability of explicitly modeling the deformational characteristics of the base-isolation 
system. This is especially important for systems that have damping characteristics that are 
amplitude-dependent, rather than velocity-dependent, since it is difficult to determine an appropriate 
value of equivalent viscous damping for these systems. 
2003 Commentary, Chapter 13 
282 
Sec. 13.2.4 of this commentary discusses other conditions which require use of the response history 
procedure. 
When response history analysis is used as the basis for design, the design displacement of the isolation 
system and design forces in elements of the structure above are to be based on the maximum of the 
results of not less than three separate analyses, each using a different pair of horizontal time histories. 
Each pair of horizontal time histories should: 
1. Be of a duration consistent with the design earthquake or the maximum considered earthquake, 
2. Incorporate near-field phenomena, as appropriate, and 
3. Have response spectra for which the square-root-of-the-sum-of-the-squares combination of the two 
horizontal components equals or exceeds 1.3 times the target spectrum at each spectral ordinate. 
The average value of seven time histories is a standard required by the nuclear industry and is 
considered appropriate for nonlinear response history analysis of seismically isolated structures. 
13.5 DESIGN REVIEW 
Review of the design and analysis of the isolation system and design review of the isolator testing 
program is mandated by the Provisions for two key reasons: 
1. The consequences of isolator failure could be catastrophic. 
2. Isolator design and fabrication technology is evolving rapidly and may be based on technologies 
unfamiliar to many design professionals. 
The Provisions requires review to be performed by a team of registered design professionals that are 
independent of the design team and other project contractors. The review team should include 
individuals with special expertise in one or more aspects of the design, analysis, and implementation of 
seismic isolation systems. 
The review team should be formed prior to the development of design criteria (including site-specific 
ground shaking criteria) and isolation system design options. Further, the review team should have full 
access to all pertinent information and the cooperation of the design team and regulatory agencies 
involved with the project. 
13.6 TESTING 
The design displacements and forces developed from the Provisions are predicated on the basis that the 
deformational characteristics of the base isolation system have been previously defined by a 
comprehensive set of tests. If a comprehensive amount of test data are not available on a system, major 
design alterations in the building may be necessary after the tests are complete. This would result from 
variations in the isolation-system properties assumed for design and those obtained by test. Therefore, it 
is advisable that prototype systems be tested during the early phases of design, if sufficient test data is 
not available on an isolation system. 
Typical force-deflection (or hysteresis) loops are shown in Figure C13.6-1; also included are the 
definitions of values used in Sec. 13.6.2. 
Seismically Isolated Structure Design Requirements 
283 
The required sequence of tests will verify experimentally: 
1. The assumed stiffness and capacity of the wind-restraining mechanism; 
2. The variation in the isolator's deformational characteristics with amplitude (and with vertical load, if 
it is a vertical load-carrying member); 
3. The variation in the isolator's deformational characteristics for a realistic number of cycles of 
loading at the design displacement; and 
4. The ability of the system to carry its maximum and minimum vertical loads at the maximum 
displacement. 
Force-deflection tests are not required if similarly sized components have been tested previously using 
the specified sequence of tests. 
Variations in effective stiffness greater than 15 percent over 3 cycles of loading at a given amplitude, or 
greater than 20 percent over the larger number of cycles at the design displacement, would be cause for 
rejection. The variations in the vertical loads required for tests of isolators which carry vertical, as well 
as lateral, load are necessary to determine possible variations in the system properties with variations in 
overturning force. The appropriate dead loads and overturning forces for the tests are defined as the 
average loads on a given type and size of isolator for determining design properties and are the absolute 
maximum and minimum loads for the stability tests. 
13.6.4 Design properties of the isolation system 
13.6.4.1 Maximum and minimum effective stiffness. The effective stiffness is determined from the 
hysteresis loops shown in Figure C13.6-1). Stiffness may vary considerably as the test amplitude 
Figure C13.6-1 The effect of stiffness on an isolation bearing. 
2003 Commentary, Chapter 13 
284 
increases but should be reasonably stable (within 15 percent) for more than 3 cycles at a given 
amplitude. 
The intent of these requirements is to ensure that the deformational properties used in design result in 
the maximum design forces and displacements. For determining design displacement, this means using 
the lowest damping and effective-stiffness values. For determining design forces, this means using the 
lowest damping value and the greatest stiffness value. 
13.6.4.2 Effective damping. The determination of equivalent viscous damping is reasonably reliable 
for systems whose damping characteristics are velocity dependent. For systems that have amplitudedependent, 
energy-dissipating mechanisms, significant problems arise in determining an equivalent 
viscous-damping value. Since it is difficult to relate velocity and amplitude-dependent phenomena, it is 
recommended that when the equivalent-viscous damping assumed for the design of amplitudedependent, 
energy-dissipating mechanisms (such as pure-sliding systems) is greater than 30 percent, 
then the design-basis force and displacement should be determined using the response history 
procedure, as discussed in Commentary Sec. 13.2.4. 
REFERENCES 
ATC. 1982. An Investigation of the Correlation Between Earthquake Ground Motion and Building 
Performance, ATC Report 10. Redwood City, California: Applied Technology Council, 
ATC. 1986. Proceedings of a Seminar and Workshop on Base Isolation and Passive Energy 
Dissipation, ATC Report 17. Redwood City, California: Applied Technology Council. 
ATC. 1993. Proceedings of Seminar on Seismic Isolation, Passive Energy Dissipation, and Active 
Control, ATC 17-1.,Redwood City, California: Applied Technology Council. 
ATC. 2002. Seminar on Response Modification Technologies for Performance-Based Seismic Design, 
ATC 17-2. Redwood City, California: Applied Technology Council. 
AASHTO. 1999. Guide Specification for Seismic Isolation Design, Washington, D.C: American 
Association of State Highway and Transportation Officials. 
ASCE. 2003. Minimum Design Loads for Buildings and Other Structures, ASCE-7-02, Reston, VA: 
American Society of Civil Engineers. 
ASCE. 1989, 1991, 1993, and 1995. Seismic Engineering: Research and Practice. New York City: 
American Society of Civil Engineers. 
Constantinou, M. C., C. W. Winters, and D. Theodossiou. 1993. Evaluation of SEAOC and UBC 
analysis procedures, Part 2: Flexible superstructure, in Proceedings of a Seminar on Seismic Isolation, 
Passive Energy Dissipation and Active Control, ATC Report 17-1. Redwood City, California: ATC. 
Constantinou, M.C., P. Tsopelas, A. Kasalanati, and E. D. Wolff. 1999. Property modification factors 
for seismic isolation bearings, Report No. MCEER-99-0012, Multidisciplinary Center for Earthquake 
Engineering Research, Buffalo, NY. 
Earthquake Engineering Research Institute. 1990. Seismic isolation: from idea to reality, Earthquake 
Spectra Journal 6:2. 
Hussain, S.M.. 1998. Testing and quality control issues for seismic isolation systems-an overview, in 
Proceedings of of the Structural Engineers World Congress, Paper T163-4, San Francisco, CA. 
International Code Council, 2003. International Building Code, Falls Creek, VA 
Kircher, C. A., B. Lashkari, R. L. Mayes, and T. E. Kelly. 1988. Evaluation of nonlinear response in 
seismically isolated buildings, in Proceedings of a Symposium on Seismic, Shock and Vibration 
Isolation, ASTHMA PVP Conference. 
Seismically Isolated Structure Design Requirements 
285 
Lashkari, B., and C. A. Kircher. 1993. Evaluation of SEAOC & UBC analysis procedures, part 1: stiff 
superstructure, in Proceedings of a Seminar on Seismic Isolation, Passive Energy Dissipation and 
Active Control, ATC Report 17-1. Redwood City, California: ATC. 
Morgan, T., A. S. Whittaker, and A. C. Thompson. 2001. Cyclic behavior of high-damping rubber 
bearings, Proceedings, Fifth World Congress on Joints, Bearings and Seismic Systems for Concrete 
Structures, American Concrete Institute, Rome, Italy. 
Skinner, R. I., W. H. Robinson, and G. H. McVerry. 1993. An Introduction to Seismic Isolation. 
Sussex, England: Wiley and Sons. 
Thompson, A., A. S. Whittaker, G. L. Fenves, and S. A. Mahin. 2000. Property modification factors for 
elastomeric bearings, Proceedings, Twelfth World Conference on Earthquake Engineering, Auckland, 
New Zealand. 

2003 Commentary, Chapter 13 
286 
This page is intentionally left blank. 
 287 
Chapter 14 Commentary 
NONBUILDING STRUCTURE DESIGN REQUIREMENTS 
14.1 GENERAL 
14.1.1 Scope. Requirements concerning nonbuilding structures were originally added to the 1994 
Provisions by the 1991-94 Provisions Update Committee (PUC) at the request of the BSSC Board of 
Direction to provide building officials with needed guidance. In recognition of the complexity, nuances, 
and importance of nonbuilding structures, the BSSC Board established 1994-97 PUC Technical 
Subcommittee 13 (TS13), Nonbuilding Structures, in 1995. The duties of TS13 were to review the 1994 
Provisions and Commentary and recommend changes for the 1997 Edition. The subcommittee comprised 
individuals possessing considerable expertise concerning various specialized nonbuilding structures and 
representing a wide variety of industries concerned with nonbuilding structures. 
Building codes traditionally have been perceived as minimum standards of care for the design of 
nonbuilding structures and building code compliance of these structures is required by building officials 
in many jurisdictions. However, requirements in the industry standards are often at odds with building 
code requirements. In some cases, the industry standards need to be altered while in other cases the 
building codes need to be modified. Registered design professionals are not always aware of the 
numerous accepted standards within an industry and may not know whether the accepted standards are 
adequate. It is hoped that Chapter 14 of the Provisions appropriately bridges the gap between building 
codes and existing industry standards. 
One of the goals of TS13 was to review and list appropriate industry standards to serve as a resource. 
These standards had to be included in the appendix. The subcommittee also has attempted to provide an 
appropriate link so that the accepted industry standards can be used with the seismic ground motions 
established in the Provisions. It should be noted that some nonbuilding structures are very similar to a 
building and can be designed employing sections of the Provisions directly whereas other nonbuilding 
structures require special analysis unique to the particular type of nonbuilding structure. 
The ultimate goal of TS13 was to provide guidance to develop requirements consistent with the intent of 
the Provisions while allowing the use of accepted industry standards. Some of the referenced standards 
are consensus documents while others are not. 
One good example of the dilemma posed by the conflicts between the Provisions and accepted design 
practice for nonbuilding structures involves steel multilegged water towers. Historically, such towers 
have performed well when properly designed in accordance with American Water Works Association 
(AWWA) standards, but these standards differ from the Provisions that tension-only rods are required and 
the connection forces are not amplified. However, industry practice requires upset rods that are preloaded 
at the time of installation, and the towers tend to perform well in earthquake areas. 
In an effort to provide the appropriate interface between the Provisions requirements for building 
structures, nonstructural components, and nonbuilding structures; TS13 recommended that nonbuilding 
structure requirements be placed in a separate chapter. The PUC agreed with this change. The 1997 
Provisions Chapter 14 now provides registered design professionals responsible for designing 
nonbuilding structures with a single point of reference. 
Note that building structures, vehicular and railroad bridges, electric power substation equipment, 
overhead power line support structures, buried pipelines and conduits, tunnels, lifeline systems, nuclear 
power plants, and dams are excluded from the scope of the nonbuilding structure requirements. The 
excluded structures are covered by other well established design criteria (e.g., electric power substation 
equipment, power line support structures, vehicular and railroad bridges), are not under the jurisdiction of 
2 003 Commentary, Chapter 14 
288 
local building officials (e.g., nuclear power plants, and dams), or require technical considerations beyond 
the scope of the Provisions (e.g., piers and wharves, buried pipelines and conduits, tunnels, and lifeline 
systems). Since many components of lifeline systems can be designed in accordance with the Provisions, 
the following information is provided to clarify why lifeline systems are excluded from the scope of the 
Provisions. 
Seismic design for a lifeline system will typically require consideration of factors that are unique to or 
particularly important to that specific system. Seismic design requirements for lifeline systems will 
typically differ from those for buildings individual structural components for the following reasons: 
1. Physical characteristics. A building consists of structural and non-structural components within a 
single site, whereas lifeline systems consist of networks of multiple and spatially distributed linked 
components (primarily non-building structures and equipment, and possibly some buildings as well.) 
2. Stakeholders. The stakeholders in the continued operation of a building after an earthquake are a 
relatively small group of building owners, tenants, and insurers. Lifeline systems provide essential 
services to a community (e.g., electric power,, communications, transportation, natural gas, water, 
wastewater, and liquid fuel). Therefore, stakeholders in the seismic performance of such systems are 
the businesses and residents of the region served by the system, business clients/vendors outside of 
the region whose continued operation will be impacted by the conditions of the businesses/residents 
within the region, and the lifeline systems owners and insurers. 
3. Performance.. Acceptable seismic performance of a building is typically measured by whether life 
safety of building occupants has been adequately protected (in accordance with minimum building 
code design provisions.) In addition, for those relatively few buildings for which performance based 
design has been considered, acceptable seismic performance will also be measured by how well postearthquake 
functionality and return-to-service requirements of the building tenants have been met. 
The ability of a lifeline system to maintain an acceptable level of service after an earthquake will 
depend, not only on the seismic performance of its various spatially dispersed components, but also 
on the redundancy and service capacity of these components (e.g., number of lanes within roadway 
elements). To the extent that a lifeline system is comprised of redundant components of sufficient 
service capacity, it can maintain an acceptable level of service to a community even if some of the 
redundant components are damaged during the earthquake. In addition, except for certain 
transportation structures (e.g., bridges and tunnels), earthquake damage to the lifeline system 
components generally do not result in direct life-safety consequences. Therefore, acceptable seismic 
performance for a lifeline system is typically based on: (a) whether the system provides an adequate 
level of service to its users after an earthquake; (b) whether economic losses related to direct damage, 
lost revenue from an inoperable system, and liability exposure are within tolerable limits; and (c) 
whether any adverse political, legal, social, administrative, or environmental consequences are 
experienced. For these reasons, acceptable seismic performance requirements for lifeline systems are 
best established through interaction with the appropriate stakeholders, including the lifeline agency, 
its customers or users, and appropriate regulatory interests. 
The definition of what constitutes a component of a lifeline system is often complicated. Components of 
utility lifeline systems are typically identical to components that might be found in industrial or 
commercial applications. A good example of this overlap are aboveground storage tanks that are 
common in large industrial or manufacturing facilities as well as water and liquid hydrocarbon 
transportation systems. Because of this similarity, a clear definition is needed to determine when design 
in accordance with the recommended approach for lifeline systems should be give preference over 
requirements in the Provisions. Three criteria are considered for determining whether the design of a 
particular nonbuilding structure can be treated as a component of a lifeline system. 
1. Spatial distribution. As noted above, lifeline systems are typically spatially-distributed systems that 
provide services considered essential to community activities and include electric power, 
communications, water, waste-water, natural gas, liquid fuel, and transportation systems. Fixed 
facilities, such as power plants, compressor stations, metering stations, are typically treated as nodes 
 Nonbuilding Structure Design Requirements 
289 
of a lifeline system and are designed in accordance with these Provisions. 
2. Definition by legal boundary. Portions of utility lifeline systems upstream of the point defining the 
legal boundary for ownership and responsibility for maintenance and repair shall be considered as 
part of a lifeline system. The physical elements of transportation lifeline systems not excluded in the 
Provisions and owned and maintained by a transportation agency are also considered part of a lifeline 
system. 
Defining lifeline system components by a legal boundary is most appropriate for utility systems that 
deliver electric power, natural gas, electric power, wastewater and some telecommunication services. 
Existing regulatory provisions commonly specify a specific interface between the portions of these 
systems that is under the control of the service provider and the portions of the system under control 
of the building or facility owner. For electric power, natural gas, and water systems, this boundary is 
typically the customers side of the meter. The other typical boundary is the property line. Those 
components under control of the service provider can be considered as part of a lifeline system. 
It is common for the design and maintenance of physical elements of transportation lifeline systems to 
fall under the jurisdiction of a governmental or government-regulated entity. Two common examples 
include state highway departments and port authorities. In such cases, the definition of a lifeline 
system by legal boundary for these situations is defined by the jurisdiction of these agencies. 
3. Definition by expertise. Historically, the primary audience of the Provisions has been the structural 
engineering community and building code organizations seeking to modify their seismic provisions. 
As a result of this focus, the Provisions are best suited for the seismic design and performance of 
individual structures. Since most new construction for lifeline systems address adding components to 
existing systems, rational design approaches should consider the overall system performance in 
design of new components and the benefits of improved seismic performance in comparison with the 
performance of the system for other natural and other hazards, such as man-made threats. The 
geographically diverse nature of lifeline systems often requires that earthquake hazards be defined by 
one or more scenario events instead of the probabilistic ground motion hazards defined in the 
Provisions. These additional considerations often require special expertise in addition to that of the 
structural engineering profession that is dominant audience for the Provisions. 
14.1.2 References 
American Concrete Institute (ACI): 
ACI 350, Environmental Engineering Concrete Structures, 2001. 
ACI 350.3 Seismic Design of Liquid-Containing Concrete Structures, 2001. 
American Society of Civil Engineers (ASCE): 
Guidelines for Seismic Evaluation and Design of Petrochemical Facilities, 1997. 
Housner, G.W. Earthquake Pressures in Fluid Containers, California Institute of Technology, 1954. 
Miller, C. D., S. W. Meier, and W. J. Czaska, Effects of Internal Pressure on Axial Compressive Strength 
of Cylinders and Cones, Structural Stability Research Council Annual Technical Meeting, June 1997. 
Rack Manufacturers Institute: 
Specification for the Design, Testing, and Utilization of Industrial Steel Storage Racks, 1997 
(Reaffirmed 2002). 
Troitsky, M.S., Tubular Steel Structures, 1990. (Troitsky) 
Wozniak, R. S., and W. W. Mitchell, Basis of Seismic Design Provisions for Welded Steel Oil Storage 
Tanks, 1978 ProceedingsRefining Dept., Vol. 57, American Petroleum Institute, Washington, D.C., 
May 9, 1978. (Wozniak) 
2 003 Commentary, Chapter 14 
290 
14.1.2.2 Other references 
While not cited directly in the Provisions or Commentary, the user may find these other references related 
to nonbuilding structures helpful. 
ACI 307, Standard Practice for the Design and Construction of Cast-In-Place Reinforced 
Concrete Chimneys, 1995. 
ANSI K61.1 Safety Requirements for the Storage and Handling of Anhydrous Ammonia, 
American National Standards Institute, 1999. 
API 2510 Design and Construction of Liquefied Petroleum Gas Installation, American 
Petroleum Institute, 19952001. 
ASCE Petro Design of Secondary Containment in Petrochemical Facilities 
(Petrochemical Energy Committee Task Report), American Society of Civil 
Engineers, 1997. 
ASME B31.8 Gas Transmission and Distribution Piping Systems, American Society of 
Mechanical Engineers, 1995. 
ASME B96.1 Welded Aluminum-Alloy Storage Tanks, American Society of Mechanical 
Engineers, 1993. 
ASME STS-1 Steel Stacks, American Society of Mechanical Engineers, 2001. 
ASTM F 1159 Standard Practice for the Design and Manufacture of Amusement Rides and 
Devices (ASTM F 1159-97a), American Society for Testing and Materials, 1997. 
ASTM C 1298 Standard Guide for Design and Construction of Brick Liners for Industrial 
Chimneys (ASTM C 1298-95), American Society for Testing and Materials, 
1995. 
DOT 49CFR193 Liquefied Natural Gas Facilities: Federal Safety Standards (Title 49CFR Part 
193), U.S. Department of Transportation, 2000. 
NFPA 30 Flammable and Combustible Liquids Code, National Fire Protection Association, 
12000. 
NFPA 58 Storage and Handling of Liquefied Petroleum Gas, National Fire Protection 
Association, 2001. 
NFPA 59 Storage and Handling of Liquefied Petroleum Gases at Utility Gas Plants, 
National Fire Protection Association, 2001. 
NFPA 59A Production, Storage and Handling of Liquefied Natural Gas (LNG), National 
Fire Protection Association, 2001. 
NCEL R-939 Ebeling, R. M., and Morrison, E. E., The Seismic Design of Waterfront Retaining 
Structures, Naval Civil Engineering Laboratory, 1993. 
NAVFAC DM-25.1 Piers and Wharves, U.S. Naval Facilities Engineering Command, 1987. 
TM 5-809-10 Seismic Design for Buildings, U.S. Army Corps of Engineers, 1992, Chapter 13 
only. 
14.1.5 Nonbuilding structures supported by other structures. This section has been developed to 
provide an appropriate link between the requirements for nonbuilding structures and those for inclusion in 
the rest of the Provisionsespecially the requirements for architectural, mechanical, and electrical 
components. 
14.2 GENERAL DESIGN REQUIREMENTS 
14.2.1 Seismic use groups and importance factors. The Importance Factors and Seismic Use Group 
classifications assigned to nonbuilding structures vary from those assigned to building structures. 
 Nonbuilding Structure Design Requirements 
291 
Buildings are designed to protect occupants inside the structure whereas nonbuilding structures are not 
normally occupied in the same sense as buildings, but need to be designed in a special manner because 
they pose a different sort of risk in regard to public safety (that is, they may contain very hazardous 
compounds or be essential components in critical lifeline systems). For example, tanks and vessels may 
contain materials that are essential for lifeline functions following a seismic event (such as fire-fighting or 
potable water), potentially harmful or hazardous to the environment or general health of the public, 
biologically lethal or toxic, or explosive or flammable (posing a threat of consequential or secondary 
damage). 
If not covered by the authority having jurisdiction, Table 14.2-1 may be used to select the importance 
factor (I). The value shall be determined by taking the larger of the value from the approved Standard or 
the value selected from Table 14.2-1. It should be noted that a single value of importance factor may not 
apply to an entire facility. For further details, refer to ASCE Petro. The use of a secondary containment 
system, when designed in accordance with an acceptable National Standard, could be considered as an 
effective means to contain hazardous substances and thus reduce the hazard classification. 
The specific definition of material hazard and what constitutes a hazard is being developed in the 
International Building Code process. The hazards will be predicated on the quantity and type of 
hazardous material. 
The importance factor is not intended for use in making economic evaluations regarding the level of 
damage, probabilities of occurrence, or cost to repair the structure. These economic decisions should be 
made by the owner and other interested parties (insurers, financiers, etc.). Nor it is intended for use for 
purposes other than that defined in this provision. 
Examples are presented below demonstrate how this table may be applied. 
Example 1. A water storage tank used to provide pressurized potable water for a process within a 
chemical plant where the tank is located away from personnel working within the facility. 
Table 14.2-1 Seismic Use Groups and Importance Factors for Nonbuilding Structures 
Seismic Use Group 
I 
II 
III 
Function 
F-I 
F-II 
F-III 
Hazard 
H-I 
H-II 
H-III 
Importance Factor 
I = 1.0 
I = 1.25 
I = 1.5 
Address each of the issues implied in the matrix: 
$ Seismic Use Group: Neither the structure nor the contents are critical, therefore use Seismic Use 
Group I. 
$ Function: The water storage tank is neither a designated ancillary structure for post-earthquake 
recovery, nor identified as an emergency back-up facilities for a Seismic Use Group III structure, 
therefore use F-I. 
$ Hazard: The contents are not hazardous, therefore use H-I. 
$ This tank has an importance factor of 1.0. 

2 003 Commentary, Chapter 14 
292 
Example 2. A steel storage rack is located in a retail store in which the customers have direct access to 
the aisles. Merchandise is stored on the upper racks. The rack is supported by a slab on grade. 
Table 14.2-1 Seismic Use Groups and Importance Factors for Nonbuilding Structures 
Seismic Use Group 
I 
II 
III 
Function 
F-I 
F-II 
F-III 
Hazard 
H-I 
H-II 
H-III 
Importance Factor 
I = 1.0 
I = 1.25 
I = 1.5 
Address each of the issues in the matrix: 
$ Seismic Use Group: Neither the structure nor the contents are critical, therefore use Seismic Use 
Group I. 
$ Function: The storage rack is neither used for post-earthquake recovery, nor required for emergency 
back-up, therefore use F-I. 
$ Hazard: The contents are not hazardous. However, its use could cause a substantial public hazard 
during an earthquake. Subject to the local authoritys jurisdiction it is H-II. 
$ According to Sec. 14.3.5.2 the importance factor for storage racks in occupancies open to the general 
public must be taken as 1.5. 
$ Use an importance factor of 1.5 for this structure. 
Example 3. A water tank is located within an office building complex to supply the fire sprinkler system. 
Table 14.2-1 Seismic Use Groups and Importance Factors for Nonbuilding Structures 
Seismic Use Group 
I 
II 
III 
Function 
F-I 
F-II 
F-III 
Hazard 
H-I 
H-II 
H-III 
Importance Factor 
I = 1.0 
I = 1.25 
I = 1.5 
Address each of the issues in the matrix: 
$ Seismic Use Group: The office building is assigned to Seismic Use Group I. 
$ Function: The water tank is required to provide water for fire fighting. However since the building is 
not a Seismic Use Group III structure, the water is used neither for post-earthquake recovery, nor for 
emergency back-up, so use F-I. 
$ Hazard: The content and its use are not hazardous to the public, therefore use H-I. 
$ Use an importance factor of 1.0 for this water structure. 

 Nonbuilding Structure Design Requirements 
293 
Example 4. A petrochemical storage tank is to be constructed within a refinery tank farm near a 
populated city neighborhood. An impoundment dike is provided to control liquid spills. 
Table 14.2-1 Seismic Use Groups and Importance Factors for Nonbuilding Structures 
Seismic Use Group 
I 
II 
III 
Function 
F-I 
F-II 
F-III 
Hazard 
H-I 
H-II 
H-III 
Importance Factor 
I = 1.0 
I = 1.25 
I = 1.5 
Address each of the issues in the matrix: 
$ Seismic Use Group: The LNG tank is assigned to Seismic Use Group III. 
$ Function: The tank is neither required to provide post-earthquake recovery nor used for emergency 
back-up for a Seismic Use Group III structure, so use F-I. 
$ Hazard: The tank contains a substantial quantity of high explosive and is near a city neighborhood. 
Despite the diking, it is considered hazardous to the public in the event of an earthquake, so use H-III. 
$ Use an importance factor of 1.5 for this structure. 
14.2.3 Design basis. The design basis for nonbuilding structures is based on either adopted references, 
approved standards, or these Provisions. It is intended that the Provisions applicable to buildings apply to 
nonbuilding structures, unless specifically noted in this Chapter. 
14.2.4 Seismic force-resisting system selection and limitations. Nonbuilding structures similar to 
buildings may be designed in accordance with either Table 4.3-1 or Table 14.2-2, including referenced 
design and detailing requirements. For convenience, Table 4.3-1 requirements are repeated in Table 14.2- 
2. 
Table 14.2-2 of the 2000 NEHRP Provisions for nonbuilding structures similar to buildings prescribed R, 
Oo
, and Cd values to be taken from Table 4.3-1, but prescribed less restrictive height limitations than 
those prescribed in Table 4.3-1. This inconsistency has been corrected. Nonbuilding structures similar to 
buildings which use the same R, Oo
, and Cd values as buildings now have the same height limits, 
restrictions and footnote exceptions as buildings. The only difference is that the footnote exceptions for 
buildings apply to metal building like systems while the exceptions for nonbuilding structures apply to 
pipe racks. In addition, selected nonbuilding structures similar to buildings have prescribed an option 
where both lower R values and less restrictive height limitations are specified. This option permits 
selected types of nonbuilding structures which have performed well in past earthquakes to be constructed 
with less restrictions in Seismic Design Categories D, E and F provided seismic detailing is used and 
design force levels are considerably higher. It should be noted that revised provisions are considerably 
more restrictive than those prescribed in Table 4.3-1. 
Nonbuilding structures not similar to buildings should be designed in accordance Table 14.2-3 
requirements, including referenced design and detailing requirements. 
Nonbuilding structures not referenced in either Table 14.2-2, Table 14.2-3, or Table 4.3-1 may be 
designed in accordance with an adopted reference, including its design and detailing requirements. 
It is not consistent with the intent of the Provisions to take design values from one table or standard and 
design and/or detailing provisions from another. 

2 003 Commentary, Chapter 14 
294 
14.2.5 Structural analysis procedure selection. Nonbuilding structures that are similar to buildings 
should be subject to the same analysis procedure limitations as building structures. 
Nonbuilding structures that are not similar to buildings should not be subject to these procedure 
limitations. However, they should be subject to any procedure limitations prescribed in specific adopted 
references. 
For nonbuilding structures supporting flexible system components, such as pipe racks, the supported 
piping and platforms are generally not regarded as rigid enough to redistribute seismic forces to the 
supporting frames. 
For nonbuilding structures supporting rigid system components, such as steam turbine generators (STGs) 
and Heat Recovery Steam Generators (HRSGs), the supported equipment, ductwork, and other 
components (depending on how they are attached to the structure) may be rigid enough to redistribute 
seismic forces to the supporting frames. Torsional effects may need to be considered in such situations. 
14.2.9 Fundamental period. The rational methods for period calculation contained in the Provisions 
were developed for building structures. If the nonbuilding structure has dynamic characteristics similar to 
those of a building, the difference in period is insignificant. If the nonbuilding structure is not similar to a 
building structure, other techniques for period calculation will be required. Some of the references for 
specific types of nonbuilding structures contain more accurate methods for period determination. 
Equations 5.2-6, 5.2-7, and 5.2-7 are not recommended because they are not relevant for the commonly 
encountered nonbuilding structures. 
14.3 NONBUILDING STRUCTURES SIMILAR TO BUILDINGS 
Nonbuilding structures exhibit behavior similar to that of building structures; however, their function and 
performance are different. Although the Provisions for buildings are used as the primary basis for design, 
this section identifies appropriate exceptions, modifications, and additions for selected nonbuilding 
structures similar to buildings. 
14.3.1 Electrical power generating facilities. Electrical power plants closely resemble building 
structures, and their performance in seismic events has been good. For reasons of mechanical 
performance, lateral drift of the structure must be limited. The lateral bracing system of choice has been 
the concentrically braced frame. The height limits on braced frames in particular can be an encumbrance 
to the design of large power generation facilities. 
14.3.3 Piers and wharves. Current industry practice recognizes the distinct differences between the two 
categories of piers and wharves described in the Provisions. The piers and wharves with public 
occupancy, described in paragraph (a) are commonly treated as the foundation for buildings or 
building-like structures, and design is performed using the Provisions. The design is likely to be under 
the jurisdiction of the local building official. 
Piers and wharves where occupancy by the general public is not a consideration, as described in 
paragraph (b), are often treated differently. In many cases, they do not fall under the jurisdiction of 
building officials, and utilize other design approaches more common to this industry. 
Economics plays a major role in the design decisions associated with these structures. These economic 
decisions may be affected not only by the wishes of the owners, but also by overlapping jurisdictional 
entities with local, regional, or state interests in commercial development. 
In the cases where the Building Officials have jurisdiction, they typically do not have experience 
analyzing pier and wharf structures. In these instances, they have come to rely on and utilize the other 
design approaches that are more common in the industry. 
Major ports and marine terminals in seismic regions of the world routinely design structures as described 
in paragraph (b). The design of these often uses a performance-based approach, with criteria and methods 
that are very different than those used for buildings, as provided in the Provisions. 
Design approaches most commonly used are generally consistent with the practices and criteria described 
 Nonbuilding Structure Design Requirements 
295 
in the following documents: 
 Working Group No. 34 of the Maritime Navigation Commission 
(PIANC/MarCom/WG34), 2001, Seismic Design Guidelines for Port Structures, A. A. 
Balkema, Lisse, Netherlands, 2001. 
 Ferritto, J., Dickenson, S., Priestley N., Werner, S., Taylor, C., Burke D., Seelig W., and 
Kelly, S., 1999, Seismic Criteria for California Marine Oil Terminals, Vol.1 and Vol.2, 
Technical Report TR-2103-SHR, Naval Facilities Engineering Service Center, Port 
Hueneme, CA. 
 Priestley, N.J.N., Frieder Siebel, Gian Michele Calvi, Seismic Design and Retrofit of 
Bridges, 1996, New York. 
 Seismic Guidelines for Ports, by the Ports Committee of the Technical Council on 
Lifeline Earthquake Engineering, ASCE, edited by Stuart D. Werner, Monograph No. 12, 
March 1998, published by ASCE, Reston, VA. 
 Marine Oil Terminal Engineering and Maintenance Standards, California State Lands 
Commission, Marine Facilities Division, May 2002. 
These alternative approaches have been developed over a period of many years by working groups within 
the industry, and consider the historical experience and performance characteristics of these structures 
that are very different than building structures. 
The main emphasis of the performance-based design approach is to provide criteria and methods that 
depend on the economic importance of a facility. Adherence to the performance criteria in the documents 
listed above is expected to provide as least as much inherent life-safety, and likely much more, than for 
buildings designed using the Provisions. However, the philosophy of these criteria is not to provide 
uniform margins of collapse for all structures. Among the reasons for the higher inherent level of lifesafety 
for these structures are the following: 
 These structures have relatively infrequent occupancy, with few working personnel and very 
low density of personnel. Most of these structures consist primarily of open area, with no 
enclosed building structures which can collapse onto personnel. Small control buildings on 
marine oil terminals or similar secondary structures are commonly designed in accordance 
with the local building code. 
 These pier or wharf structures are typically constructed of reinforced concrete, prestressed 
concrete, and/or steel and are highly redundant due to the large number of piles supporting a 
single wharf deck unit. Tests done for the Port of Los Angeles at the University of California 
at San Diego have shown that very high ductilities (10 or more) can be achieved in the design 
of these structures using practices currently used in California ports. 
 Container cranes, loading arms, and other major structures or equipment on the piers or 
wharves are specifically designed not to collapse in an earthquake. Typically, additional 
piles and structural members are incorporated into the wharf or pier specifically to support 
that item. 
 Experience has shown that seismic failure of wharf structures in zones of strong seismicity 
is indicated not by collapse, but by economically unrepairable deformations of the piles. The 
wharf deck generally remains level or slightly tilting but shifted out of position. Complete 
failure that could cause life-safety concerns has not been known to ever occur historically due 
to earthquake loading. 

2 003 Commentary, Chapter 14 
296 
 The performance-based criteria of the listed documents include repairability of the structure. 
This service level is much more stringent than collapse prevention and would provide a 
greater margin for life-safety. 
 Lateral load design of these structures is often governed by other marine loading conditions, 
such as mooring or berthing. 
14.3.4 Pipe racks. Free standing pipe racks supported at or below grade with framing systems that are 
similar in configuration to building systems should be designed to satisfy the force requirements of 
Sec. 5.2. Single column pipe racks that resist lateral loads should be designed as inverted pendulums. 
See ASCE Petro. 
14.3.5 Steel storage racks. This section is intended to assure comparable results from the use of the 
RMI Specification, the NEHRP Provisions, and the IBC code approaches to rack structural design. 
For many years the RMI has been working with the various committees of the model code organizations 
and with the Building Seismic Safety Council and its Technical Subcommittees to create seismic design 
provisions particularly applicable to steel storage rack structures. The 1997 RMI Specification is seen to 
be in concert with the needs, provisions, and design intent of the building codes and those who use and 
promulgate them, as well as those who engineer, manufacture, install, operate, use, and maintain rack 
structures. The RMI Specification, now including detailed seismic provisions, is essentially selfsufficient. 
The changes proposed here are compatible and coordinated with those in the 2000 International Building 
Code. 
14.3.5.2 Importance factor. Until recently, storage racks were primarily installed in low-occupancy 
warehouses. With the recent proliferation of warehouse-type retail stores, it has been judged necessary to 
address the relatively greater seismic risk that storage racks may pose to the general public, compared to 
more conventional retail environments. Under normal operating conditions, retail stores have a far higher 
occupancy load than an ordinary warehouse of a reasonable size. Failure of a storage rack system in the 
retail environment is much more likely to cause personal injury than a similar failure in a storage 
warehouse. Therefore, to provide an appropriate level of additional safety in areas open to the public, 
Sec 14.3.5.2 now requires that storage racks in occupancies open to the general public be designed with 
an importance factor equal to 1.50. Storage rack contents, while beyond the scope of the Provisions, pose 
a potentially serious threat to life should they fall from the shelves in an earthquake. Restraints should be 
provided to prevent the contents of rack shelving open to the general public from falling in strong ground 
shaking. 
14.4 NONBUILDING STRUCTURES NOT SIMILAR TO BUILDINGS 
Nonbuilding structures not similar to buildings exhibit behavior markedly different from that of building 
structures. Most of these types of structures have adopted references that address their unique structural 
performance and behavior. The ground motion in the Provisions requires appropriate translation to allow 
use with industry standards. Such translation is provided in this section. 
14.4.2 Earth retaining structures. In order to properly develop and implement methodologies for the 
design of earth retaining structures, it is essential to know and understand the nature of the applied loads. 
Concerns have been raised concerning the design of nonyielding walls and yielding walls for bending, 
overturning, sliding, etc., taking into account the varying soil types, importance, and site seismicity. See 
Sec. 7.5.1 in the Commentary. 
14.4.3 Stacks and chimneys. The design of stacks and chimneys to resist natural hazards is generally 
governed by wind design considerations. The exceptions to this general rule involve locations with high 
seismicity, stacks and chimneys with large elevated masses, and stacks and chimneys with unusual 
geometries. It is prudent to evaluate the effect of seismic loads in all but those areas with the lowest 
seismicity. Although not specifically required, it is recommended that the special seismic details required 
elsewhere in the Provisions be evaluated for applicability to stacks and chimneys. 
 Nonbuilding Structure Design Requirements 
297 
Guyed steel stacks and chimneys are generally light weight. As a result, the design loads due to natural 
hazards are generally governed by wind. On occasion, large flares or other elevated masses located near 
the top may require an in-depth seismic analysis. Although Chapter 6 of Troitsky does not specifically 
address seismic loading, it remains an applicable methodology for resolution of seismic forces that are 
defined in these Provisions. 
14.4.7 Tanks and vessels. Methods of seismic design of tanks, currently adopted by a number of 
industry standards, have evolved from earlier analytical work by Jacobsen, Housner, Veletsos, Haroun, 
and others. The procedures used to design flat bottom storage tanks and liquid containers is based on the 
work of Housner, Wozniak, and Mitchell. The standards for tanks and vessels have specific requirements 
to safeguard against catastrophic failure of the primary structure based on observed behavior in seismic 
events since the 1930s. Other methods of analysis using flexible shell models have been proposed but are 
presently beyond the scope of these Provisions. 
These methods entail three fundamental steps: 
1. The dynamic modeling of the structure and its contents. When a liquid-filled tank is subjected to 
ground acceleration, the lower portion of the contained liquid, identified as the impulsive component 
of mass WI, acts as if it were a solid mass rigidly attached to the tank wall. As this mass accelerates, it 
exerts a horizontal force, PI, against the wall that is directly proportional to the maximum acceleration 
of the tank base. This force is superimposed on the inertia force of the accelerating wall itself, Pw. 
Under the influence of the same ground acceleration, the upper portion of the contained liquid 
responds as if it were a solid mass flexibly attached to the tank wall. This portion, which oscillates at 
its own natural frequency, is identified as the convective component Wc, and exerts a force Pc on the 
wall. The convective component oscillations are characterized by the phenomenon of sloshing 
whereby the liquid surface rises above the static level on one side of the tank, and drops below that 
level on the other. 
2. The determination of the frequency of vibration, wI, of the tank structure and the impulsive 
component; and the natural frequency of oscillation (sloshing), wc, of the convective component. 
3. The selection of the design response spectrum. The response spectrum may be site-specific or it may 
be constructed deterministically on the basis of seismic coefficients given in national codes and 
standards. Once the design response spectrum is constructed, the spectral accelerations 
corresponding to wI and wc are obtained and are used to calculate the dynamic forces PI, Pw, and Pc. 
Detailed guidelines for the seismic design of circular tanks, incorporating these concepts to varying 
degrees, have been the province of at least four industry standards: AWWA D100 for welded steel tanks 
(since 1964); API 650 for petroleum storage tanks; AWWA D110 for prestressed, wire-wrapped tanks 
(since 1986); and AWWA D115 for prestressed concrete tanks stressed with tendons (since 1995). In 
addition, API 650 and API 620 contain provisions for petroleum, petrochemical, and cryogenic storage 
tanks. The detail and rigor of analysis employed by these standards have evolved from a semi-static 
approach in the early editions to a more rigorous approach at the present, reflecting the need to factor in 
the dynamic properties of these structures. 
The requirements in Sec 14.4.7 are intended to link the latest procedures for determining design level 
seismic loads with the allowable stress design procedures based on the methods in these Provisions. 
These requirements, which in many cases identify specific substitutions to be made in the design 
equations of the national standards, will assist users of the Provisions in making consistent 
interpretations. 
ACI has published a document, ACI 350.3-01 titled Seismic Design of Liquid-Containing Concrete 
Structures. This document, which covers all types of concrete tanks (prestressed and non-prestressed, 
circular and rectilinear), has provisions made consistent with the seismic guidelines of the 2000 Provisions. 
This ACI document serves as both a practical how-to loading reference and a guide to supplement 
application of Chapter 21 Special Provisions for Seismic Design of ACI 318. 
2 003 Commentary, Chapter 14 
298 
14.4.7.1 Design basis. Two important tasks of TS 13 were (a) to partially expand the coverage of 
nonbuilding structures in the Provisions; and (b) to provide comprehensive cross-references to all the 
applicable industry standards. It is hoped that this endeavor will bring about a standardization and 
consistency of design practices for the benefit of both the practicing engineer and the public at large. 
In the case of the seismic design of nonbuilding structures, standardization requires adjustments to 
industry standards to minimize existing inconsistencies among them. However, the standardization 
process should recognize that structures designed and built over the years in accordance with industry 
standards have performed well in earthquakes of varying severity. 
Of the inconsistencies among industry standards, the ones most important to seismic design relate to the 
base shear equation. The traditional base shear takes the following form: 
w 
V ZISCW 
R 
= 
An examination of those terms as used in the different references reveals the following: 
- ZS: The seismic zone coefficient, Z, has been rather consistent among all the standards by virtue of 
the fact that it has traditionally been obtained from the seismic zone designations and maps in the 
model building codes. 
On the other hand, the soil profile coefficient, S, does vary from one standard to another. In some 
standards these two terms are combined. 
S I: The importance factor, I, has also varied from one standard to another, but this variation is 
unavoidable and understandable owing to the multitude of uses and degrees of importance of liquidcontaining 
structures. 
S C: The coefficient C represents the dynamic amplification factor that defines the shape of the design 
response spectrum for any given maximum ground acceleration. Since coefficient C is primarily a 
function of the frequency of vibration, inconsistencies in its derivation from one standard to another 
stem from at least two sources: differences in the equations for the determination of the natural 
frequency of vibration, and differences in the equation for the coefficient itself. (For example, for the 
shell/impulsive liquid component of lateral force, the steel tank standards use a constant design 
spectral acceleration (namely, a constant C) that is independent of the impulsive period T.) In 
addition, the value of C will vary depending on the damping ratio assumed for the vibrating structure 
(usually between 2 percent and 7 percent of critical). 
Where a site-specific response spectrum is available, calculation of the coefficient C is not necessary 
B except in the case of the convective component (coefficient Cc) which is assumed to oscillate with 
0.5 percent of critical damping, and whose period of oscillation is usually high (greater than 2.5 sec). 
Since site-specific spectra are usually constructed for high damping values (3 percent to 7 percent of 
critical); and since the site-specific spectral profile may not be well-defined in the high-period range, 
an equation for Cc applicable to a 0.5 percent damping ratio is necessary in order to calculate the 
convective component of the seismic force. 
S Rw: The response modification factor, Rw, is perhaps the most difficult to quantify, for a number of 
reasons. While Rw is a compound coefficient that is supposed to reflect the ductility, energydissipating 
capacity, and redundancy of the structure, it is also influenced by serviceability 
considerations, particularly in the case of liquid-containing structures. 
In the Provisions the base shear equation for most structures has been reduced to V = CsW, where the 
seismic response coefficient, Cs, replaces the product 
w 
ZSC 
R 
. Cs is determined from the design spectral 
response acceleration parameters SDS and SD1 (at short periods and at a period of 1 sec, respectively) 
which, in turn, are obtained from the mapped MCE spectral accelerations Ss and S1 obtained from the 
 Nonbuilding Structure Design Requirements 
299 
seismic maps. As in the case of the prevailing industry standards, where a site-specific response spectrum 
is available, Cs is replaced by the actual spectral values of that spectrum. 
As part of its task, TS 13 has introduced a number of provisions, in the form of bridging equations, each 
designed to provide a means of properly applying the design criteria of a particular industry standard in 
the context of these Provisions. These provisions are outlined below and are identified with particular 
types of liquid-containing structures and the corresponding standards. Underlying all these provisions is 
the understanding that the calculation of the periods of vibration of the impulsive and convective 
components is left up to the industry standards. Defining the detailed resistance and allowable stresses of 
the structural elements for each industry structure has also been left to the approved standard except in 
instances where additional information has led to additional requirements. 
It is intended that, as the relevant national standards are updated to conform to these Provisions, the 
bridging equations of Sec. 14.4.7.6, 14.4.7.7, and 14.4.7.9 will be eliminated. 
14.4.7.2 Strength and ductility. As is the case for building structures, ductility and redundancy in the 
lateral support systems for tanks and vessels are desirable and necessary for good seismic performance. 
Tanks and vessels are not highly redundant structural systems and, therefore, ductile materials and welldesigned 
connection details are needed to increase the capacity of the vessel to absorb more energy 
without failure. The critical performance of many tanks and vessels is governed by shell stability 
requirements rather than by yielding of the structural elements. For example, contrary to building 
structures, ductile stretching of the anchor bolts is a desirable energy absorption component when tanks 
and vessels are anchored. The performance of cross-braced towers is highly dependent on the ability of 
the horizontal compression struts and connection details to fully develop the tension yielding in the rods. 
In such cases, it is also important to assure that the rods stretch rather than fail prematurely in the threaded 
portion of the connection and that the connection of the rod to the column does not fail prior to yielding 
of the rod. 
14.4.7.3 Flexibility of piping attachments. The performance of piping connections under seismic 
deformations is one of the primary weaknesses observed in recent seismic events. Tank leakage and 
damage occurs when the piping connections cannot accommodate the movements the tank experiences 
during the a seismic event. Unlike the connection details used by many piping designers, which 
connections impart mechanical loading to the tank shell, piping systems in seismic areas should be 
designed in such a manner as to impose only negligible mechanical loads on the tank connection for the 
values shown in Table 14.4-1. 
In addition, interconnected equipment, walkways, and bridging between multiple tanks must be designed 
to resist the loads and displacements imposed by seismic forces. Unless multiple tanks are founded on a 
single rigid foundation, walkways, piping, bridges, and other connecting structures must be designed to 
allow for the calculated differential movements between connected structures due to seismic loading 
assuming the tanks and vessels respond out of phase. 
14.4.7.4 Anchorage. Many steel tanks can be designed without anchors by using the annular plate 
procedures given in the national standards. Tanks that must be anchored because of overturning potential 
could be susceptible to shell tearing if not properly designed. Ideally, the proper anchorage design will 
provide both a shell attachment and embedment detail that will yield the bolt without tearing the shell or 
pulling the bolt out the foundation. Properly designed anchored tanks retain greater reserve strength to 
resist seismic overload than do unanchored tanks. 
Premature failure of anchor bolts has been observed where the bolt and attachment are not properly 
aligned (that is, the anchor nut or washer does not bear evenly on the attachment). Additional bending 
stresses in threaded areas may cause the anchor to fail before yielding. 
14.4.7.5 Ground-supported storage tanks for liquids 
14.4.7.5.1 Seismic forces. The response of ground storage tanks to earthquakes is well documented by 
Housner, Mitchell and Wozniak, Veletsos, and others. Unlike building structures, the structural response 
2 003 Commentary, Chapter 14 
300 
is strongly influenced by the fluid-structure interaction. Fluid-structure interaction forces are categorized 
as sloshing (convective mass) and rigid (impulsive mass) forces. The proportion of these forces depends 
on the geometry (height-to-diameter ratio) of the tank. API 650, API 620, AWWA D100, AWWA D110, 
AWWA D115, and ACI 350.3 provide the necessary data to determine the relative masses and moments 
for each of these contributions. 
The Provisions stipulate that these structures shall be designed in accordance with the prevailing 
approved industry standards, with the exception of the height of the sloshing wave, ds, which is to be 
calculated using Eq. 14.4-9 of these Provisions. 
d s = 0.5DISac 
This equation utilizes a spectral response coefficient 1.5 D1 
ac 
c
S S 
T 
= for Tc < 4.0 sec., and 1 
2 
6 D 
ac 
c 
S S 
T 
= for 
Tc > 4.0 sec. The first definition of Sa represents the constant-velocity region of the response spectrum 
and the second the constant-displacement region of the response spectrum, both at 0.5 percent damping. 
In practical terms, the latter is the more commonly used definition since most tanks have a fundamental 
period of liquid oscillation (sloshing wave period) greater than 4.0 sec. 
Small diameter tanks and vessels are more susceptible to overturning and vertical buckling. As a general 
rule, the greater the ratio of H/D, the lower the resistance is to vertical buckling. When H/D > 2, the 
overturning begins to approach rigid mass behavior (the sloshing mass is small). Large diameter tanks 
may be governed by additional hydrodynamic hoop stresses in the middle regions of the shell. 
The impulsive period (the natural period of the tank components and the impulsive component of the 
liquid) is typically in the 0.25 to 0.6 second range. Many methods are available for calculating the 
impulsive period. The Veletsos flexible-shell method is commonly used by many tank designers. (For 
example, see Seismic Effects in Flexible Liquid Storage Tanks by A. S. Veletsos.) 
14.4.7.5.2 Distribution of hydrodynamic and inertia forces. Most of the methods contained in the 
industry standards for tanks define reaction loads at the base of the shell and foundation interface. Many 
of the standards do not give specific guidance for determining the distribution of the loads on the shell as 
a function of height. The design professional may find the additional information contained in ACI 350.3 
helpful. 
The overturning moment at the base of the shell as defined in the industry standards is only the portion of 
the moment that is transferred to the shell. It is important for the design professional to realize that the 
total overturning moment must also include the variation in bottom pressure. This is important when 
designing pile caps, slabs, or other support elements that must resist the total overturning moment. See 
Wozniak or TID 7024 for further information. 
14.4.7.5.3 Freeboard. Performance of ground storage tanks in past earthquakes has indicated that 
sloshing of the contents can cause leakage and damage to the roof and internal components. While the 
effect of sloshing often involves only the cost and inconvenience of making repairs, rather than 
catastrophic failure, even this limited damage can be prevented or significantly mitigated when the 
following items are considered: 
1. Effective masses and hydro-dynamic forces in the container. 
2. Impulsive and pressure loads at 
a. Sloshing zone (that is, the upper shell and edge of the roof system), 
b. Internal supports (roof support columns, tray-supports, etc.), and 
c. Equipment (distribution rings, access tubes, pump wells, risers, etc.). 
3. Freeboard (which depends on the sloshing wave height). 
A minimum freeboard of 0.7ds is recommended for economic considerations but is not required. 
 Nonbuilding Structure Design Requirements 
301 
Tanks and vessels storing biologically or environmentally benign materials do not typically require 
freeboard to protect the public health and safety. However, providing freeboard in areas of frequent 
seismic occurrence for vessels normally operated at or near top capacity may lessen damage (and the cost 
of subsequent repairs) to the roof and upper container. 
The estimate given in the Provision Sec. 14.4.7.5.3 is based on the seismic design event as defined by the 
Provisions. Users of the Provisions may estimate slosh heights different from those recommended in the 
national standards. 
If sloshing is restricted because the freeboard provided is less than the computed sloshing height, ds, the 
sloshing liquid will impinge on the roof in the vicinity of the roof-to-wall joint, subjecting it to a 
hydrodynamic force. This force may be approximated by considering the sloshing wave as a hypothetical 
static liquid column having a height, ds. The pressure exerted on any point along the roof at a distance ys 
above the at-rest surface of the stored liquid, may be assumed equal to the hydrostatic pressure exerted by the 
hypothetical liquid column at a distance ds  ys from the top of that column 
Another effect of a less-than-full freeboard is that the restricted convective (sloshing) mass converts into an 
impulsive mass thus increasing the impulsive forces. This effect should be taken account in the tank design. 
Preferably, sufficient freeboard should be provided whenever possible to accommodate the full sloshing 
height. 
14.4.7.5.6 Sliding resistance. Steel ground-supported tanks full of product have not been found to slide 
off foundations. A few unanchored, empty tanks have moved laterally during earthquake ground shaking. 
In most cases, these tanks may be returned to their proper locations. Resistance to sliding is obtained 
from the frictional resistance between the steel bottom and the sand cushion on which bottoms are placed. 
Because tank bottoms usually are crowned upward toward the tank center and are constructed of 
overlapping, fillet-welded, individual steel plates (resulting in a rough bottom), it is reasonably 
conservative to take the ultimate coefficient of friction as 0.70 (U.S. Nuclear Regulatory Commission, 
1989, pg. A-50) and, therefore, a value of tan 30o (= 0.577) is used. The vertical weight of the tank and 
contents as reduced by the component of vertical acceleration provides the net vertical load. An 
orthogonal combination of vertical and horizontal seismic forces following the procedure in Sec. 5.2 may 
be used. 
14.4.7.5.7 Local shear transfer. The transfer of seismic shear from the roof to the shell and from the 
shell to the base is accomplished by a combination of membrane shear and radial shear in the wall of the 
tank. For steel tanks, the radial shear is very small and is usually neglected; thus, the shear is assumed to 
be carried totally by membrane shear. For concrete walls and shells, which have a greater radial shear 
stiffness, the shear transfer may be shared. The user is referred to the ACI 350 commentary for further 
discussion. 
14.4.7.5.8 Pressure stability. Internal pressure may increase the critical buckling capacity of a shell. 
Provision to include pressure stability in determining the buckling resistance of the shell for overturning 
loads is included in AWWA D100. Recent testing on conical and cylindrical shells with internal pressure 
yielded a design methodology for resisting permanent loads in addition to temporary wind and seismic 
loads. See Miller et al., 1997. 
14.4.7.5.9 Shell support. Anchored steel tanks should be shimmed and grouted to provide proper 
support for the shell and to reduce impact on the anchor bolts under reversible loads. The high bearing 
pressures on the toe of the tank shell may cause inelastic deformations in compressible material (such as 
fiberboard), creating a gap between the anchor and the attachment. As the load reverses, the bolt is no 
longer snug and an impact of the attachment on the anchor can occur. Grout is a structural element and 
should be installed and inspected as if it is an important part of the vertical- and lateral-force-resisting 
system. 
14.4.7.5.10 Repair, alteration, or reconstruction. During their service life, storage tanks are frequently 
repaired, modified or relocated. Repairs or often related to corrosion, improper operation, or overload 
2 003 Commentary, Chapter 14 
302 
from wind or seismic events. Modifications are made for changes in service, updates to safety equipment 
for changing regulations, installation of additional process piping connections. It is imperative these 
repairs and modifications are properly designed and implemented to maintain the structural integrity of 
the tank or vessel for seismic loads as well as the design operating loads. 
The petroleum steel tank industry has developed specific guidelines in API 653 that are statutory 
requirements in some states. It is the intent of TS 13 that the provisions of API 653 also be applied to 
other liquid storage tanks (water, wastewater, chemical, etc.) as it relates to repairs, modifications or 
relocation that affects the pressure boundary or lateral force resisting system of the tank or vessel. 
14.4.7.6 Water and water treatment structures 
14.4.7.6.1 Welded steel. The AWWA design requirements for ground-supported steel water storage 
structures are based on an allowable stress method that utilizes an effective mass procedure considering 
two response modes for the tank and its contents: 
1. The high-frequency amplified response to seismic motion of the tank shell, roof, and impulsive mass 
(that portion of liquid content of the tank that moves in unison with the shell), and 
2. The low-frequency amplified response of the convective mass (that portion of the liquid contents in 
the fundamental sloshing mode). 
The two-part AWWA equation incorporates the above modes, appropriate damping, site amplification, 
allowable stress response modification, and zone coefficients. In practice, the typical ground storage tank 
and impulsive contents will have a natural period, T, of 0.1 to 0.3 sec. The sloshing period typically will 
be greater than 1 sec (usually 3 to 5 seconds depending on tank geometry). Thus, the substitution in the 
Provisions uses a short- and long-period response as it applies to the appropriate constituent term in the 
AWWA equations. 
14.4.7.6.2 Bolted steel. The AWWA Steel Tank Committee is responsible for the content of both the 
AWWA D100 and D103 and have established equivalent load and design criteria for earthquake design of 
welded and bolted steel tanks. 
14.4.7.7 Petrochemical and industrial liquids 
14.4.7.7.1 Welded steel. The American Petroleum Institute (API) also uses an allowable stress design 
procedure and the API equation has incorporated an Rw factor into the equations directly. 
The most common damage to tanks observed during past earthquakes include: 
$ Buckling of the tank shell near the base due to excessive axial membrane forces. This buckling 
damage is usually evident as elephant foot buckles a short distance above the base, or as diamond 
shaped buckles in the lower ring. Buckling of the upper ring has also been observed. 
$ Damage to the roof due to impingement on the underside of the roof of sloshing liquid with 
insufficient freeboard. 
$ Failure of piping or other attachments that are overly restrained.. 
$ Foundation failures. 
The performance of floating roofs during earthquakes has been good, with damage usually confined to the 
rim seals, gage poles, and ladders. Similarly the performance of open tops with top wind girder stiffeners 
designed per API 650 has been good. 
14.4.7.9 Elevated tanks for liquids and granular materials. There are three basic lateral-load resisting 
systems for elevated water tanks that are defined by their support structure. Multi-leg braced steel tanks 
(trussed towers), small diameter single-pedestal steel tanks (cantilever columns), and large diameter 
single-pedestal tanks of steel or concrete construction (load-bearing shear walls). Unbraced multi-leg 
tanks are not commonly built. Behavior, redundancy, and resistance to overload of these types of tanks 
are not the same. Multi-leg and small diameter pedestal have higher fundamental periods (typically over 
2-sec) than the shear wall type tanks (typically under 2-sec). Lateral load failure mechanism is usually by 
 Nonbuilding Structure Design Requirements 
303 
bracing failure for multi-leg tanks, compression buckling of small diameter steel tanks, compression or 
shear buckling of large diameter steel tanks, and shear failure of large diameter concrete tanks. In order 
to utilize the full strength of these structures adequate connection, welding, and reinforcement details 
must be provided. The R-factor used with elevated tanks is typically less than that for comparable lateral 
load-resisting systems for other purposes in order to provide a greater margin of safety. 
14.4.7.9.3 Transfer of lateral forces into support tower. The lateral transfer of load for tanks and 
vessels siting on grillage or support beams should consider the relative stiffness of the support beams and 
the shear transfer at the base of the shell, which is not typically uniform around the base of the tank. In 
addition, when tanks and vessels are supported on discrete points on grillage or beams, it is common for 
the vertical loads to vary due to settlements or variations in construction. This variation in load should be 
considered when analyzing the combined vertical and horizontal loads. 
14.4.7.9.4 Evaluation of structures sensitive to buckling failure. Nonbuilding structures that have low 
or negligible structural redundancy for lateral loads need to be evaluated for a critical level of 
performance to provide sufficient margin against premature failure. Reserve strength for loads beyond 
the design loads can be limited. Tanks and vessels supported on shell skirts or pedestals that are governed 
by buckling are examples of structures that need to be evaluated at this critical condition. Such structures 
include single pedestal water towers, process vessels, and other single member towers. 
The additional evaluation is based on a scaled maximum considered earthquake. This critical earthquake 
acceleration is defined as the design spectral response acceleration, Sa, which includes site factors. The 
I/R coefficient is taken as 1.0 for this critical check. The structural capacity of the shell is taken as the 
critical buckling strength (that is, the factor of safety is 1.0). Vertical or orthogonal earthquake 
combination need not be made for this critical evaluation since the probability of critical peak values 
occurring simultaneously is very low. 
14.4.7.9.6 Concrete pedestal (composite) tanks. A composite elevated water-storage tank is a structure 
comprising a welded steel tank for watertight containment, a single pedestal concrete support structure, 
foundation, and accessories. Lateral load-resisting system is that of a load-bearing concrete shear wall. 
Seismic provisions in ATC 371R-98 are based on ASCE 7-95, which used NEHRP 1994 as the source 
document. Seismic provisions in the proposed AWWA standard being prepared by committee D170 are 
based on ASCE 7-98, which used NEHRP 1997 as the source document. 

2 003 Commentary, Chapter 14 
304 
This page is intentionally left blank. 
 305 
Appendix to Chapter 14 
OTHER NONBUILDING STRUCTURES 
PREFACE: The following sections were originally intended to be part of the Nonbuilding 
Structures Chapter of this Commentary. The Provisions Update Committee felt that given the 
complexity of the issues, the varied nature of the resource documents, and the lack of 
supporting consensus resource documents, time did not allow a sufficient review of the 
proposed sections required for inclusion into the main body of the chapter. 
The Nonbuilding Structures Technical Subcommittee, however, expressed that what is 
presented herein represents the current industry accepted design practice within the 
engineering community that specializes in these types of nonbuilding structures. 
The Commentary sections are included here so that the design community specializing in 
these nonbuilding structures can have the opportunity to gain a familiarity with the concepts, 
update their standards, and send comments on this appendix to the BSSC. 
It is hoped that the various consensus design standards will be updated to include the design and 
construction methodology presented in this Appendix. It is also hoped that industry standards 
that are currently not consensus documents will endeavor to move their standards through the 
consensus process facilitating building code inclusion. 
A14.1 GENERAL 
Agrawal P. K., and J. M. Kramer, Analysis of Transmission Structures and Substation Structures and 
Equipment for Seismic Loading, Sargent & Lundy Transmission and Substation Conference, December 
2, 1976. (Agrawal) 
American Society of Civil Engineers (ASCE): 
ANSI/ASCE 10, Design of Latticed Transmission Structures, 1997. (ASCE 10) 
ASCE Manual 72, Tubular Pole Design Standard, 1991 (ASCE 72). 
ASCE Manual 74, Guidelines for Electrical Transmission Line Structural Loading, 2000. (ASCE 
74). 
ASCE 7, Minimum Design Loads for Buildings and Other Structures, 1995. (ASCE 7). 
ASCE Manual 91, The Design of Guyed Electrical Transmission Structures, 1997. (ASCE 91) 
Substation Structure Design Guide, 2000. (ASCE Substation) 
Li, H.-N., S. Wang, M. Lu, and Q. Wang, Aseismic Calculations for Transmission Towers, ASCE 
Technical Council on Lifeline Earthquake Engineering, Monograph No. 4, August 1991. (ASCE 
Li) 
Steinhardt, O. W., Low Cost Seismic Strengthening of Power Systems, Journal of The Technical 
Councils of ASCE, April 1981. (ASCE Steinhardt) 
Amiri, G. G. and G. G. McClure, Seismic Response to Tall Guyed Telecommunication Towers, Paper 
No. 1982, Eleventh World Conference on Earthquake Engineering, Elsevier Science Ltd., 1996. 
(Amiri) 
Australian Standards: 
Australian Standard 3995, Standard Design of Steel Lattice Towers and Masts, 1994. (AS 3995) 
Canadian Standards Association (CSA): 
2 003 Commentary, Appendix to Chapter 14 
306 
Antennas, Towers, and Masts, 1994. (CSA S37) 
Earthquake Engineering Research Institute (EERI): 
Li, H.-N., L. E. Suarez, and M. P. Singh, Seismic Effects on High-Voltage Transmission Tower 
and Cable Systems, Fifth U.S. National Conference on Earthquake Engineering, 1994. (EERI Li) 
Federal Emergency Management Agency (FEMA): 
Earthquake Resistant Construction of Electric Transmission and Telecommunication Facilities 
Serving the Federal Government, FEMA Report No. 202, September 1990. (FEMA 202) 
Galvez, C. A., and G. G. McClure, A Simplified Method for Aseismic Design of Self-Supporting 
Latticed Telecommunication Towers, Seventh Canadian Conference on Earthquake Engineering, 
Montreal, 1995. (Galvez) 
Institute of Electrical and Electronics Engineers (IEEE): 
National Electrical Safety Code, ANSI C2, New Jersey, 1997. (NESC) 
IEEE Standard 693, Recommended Practices for Seismic Design of Substations, Power Engineering 
Society, Piscataway, New Jersey, 1997 (IEEE 693). 
IEEE Standard 751, Trial-Use Design Guide for Wood Transmission Structures, Power Engineering 
Society, Piscataway, New Jersey, 1991. (IEEE 751) 
Long, L.W., Analysis of Seismic Effects on Transmission Structures, IEEE Paper T 73 326-6, April 
1973. (IEEE Long). 
Lum, W. B., N. N. Nielson, R. Koyanagi, and A. N. L. Chui, Damage Survey of the Kasiki, Hawaii 
Earthquake of November 16, 1993, Earthquake Spectra, November 1984. (Lum) 
Lyver, T. D., W. H. Mueller, and L. Kempner, Jr., Response Modification Factor, Rw, for Transmission 
Towers, Research Report, Portland State University, Portland, Oregon, 1996. (Lyver) 
National Center for Earthquake Engineering Research (NCEER): 
The Hanshin-Awaji Earthquake of January 17, 1995CPerformance of Lifelines, National Center for 
Earthquake Engineering Research, Technical Report NCEER-95-0015, State University of New 
York at Buffalo, November 3, 1995. (NCEER 95-0015) 
Rural Electrical Administration (REA): 
Bulletin 1724E-200, Design Manual for High Voltage Transmission Lines, 1992. (REA 1724). 
Bulletin 65-1, Design Guide for Rural Substations, 1978 (REA 65-1). 
Bulletin 160-2, Mechanical Design Manual for Overhead Distribution Lines, 1982. (REA 160) 
Telecommunications Industry Association (TIA): 
TIA/EIA 222F, Structural Standards for Steel Antenna Towers and Antenna Supporting Structures, 
1996. (TIA 222) 
A14.2.1 Buried Structures. This section was placed in the Appendix to Chapter 14 for the following 
reasons: 
1. The material may serve as a starting point for continued development. 
2. The comments stimulated by consideration of this section will provide valuable input so that this 
section may be further developed and then incorporated in the Provisions in the future. 
3. It was determined by TS 13 and the Provisions Update Committee that it would be premature to 
incorporate this section into the Provisions for the 2000 edition. 
4. Accepted industry standards are in the process of incorporating seismic design methodology 
reflecting the Provisions. 
Other Nonbuilding Structures 
307 
It is not the intent of the Provisions Update Committee to discourage incorporation of this section into a 
building code or to minimize the importance of this section. Placing this section in the appendix 
indicates only that this section requires further development. 
Seismic forces on buried structures may include forces due to: soil displacement, seismic lateral earth 
pressure, buoyant forces related to liquefaction, permanent ground displacements from slope instability, 
lateral spread movement, fault movement, or dynamic ground displacement caused by dynamic strains 
from wave propagation. Identification of appropriate seismic loading conditions is dependent upon 
subsurface soil conditions and the configuration of the buried structure. Conditions related to 
permanent ground movement can often be avoided by careful site selection for isolated buried structures 
such as tanks and vaults. Relocation is often impractical for long buried structures such as tunnels and 
pipelines. 
Wave propagation strains are a significant seismic force condition for buried structures if local site 
conditions (for instance, deep surface soil deposits with low shear wave velocities) can support the 
propagation of large amplitude seismic waves. Wave propagation strains tend to be most pronounced at 
the junctions of dissimilar buried structures (such as a pipeline connecting with a building) or at the 
interfaces of different geologic materials (such as a pipeline passing from rock to soft soil). 
Loading conditions related to liquefaction require detailed subsurface information that can be used to 
assess the potential for liquefaction and, for long buried structures, the length of structure exposed to 
liquefaction effects. In addition, the assessment of liquefaction requires specifying an earthquake 
magnitude that is consistent with the definition of ground shaking. It is recommended that one refer to 
Chapter 7 of this Commentary for additional guidance in determining liquefaction potential and seismic 
magnitude. Providing detailed structural design procedures in this area is beyond the scope of this 
document. 
Loading conditions related to lateral spread movement and slope instability can be defined in terms of 
lateral soil pressures or prescribed ground displacements. In both cases, sufficient subsurface 
investigation in the vicinity of the buried structure is necessary to estimate the amount of movement, the 
direction of movement relative to the buried structure, and the portion of the buried structure exposed to 
the loading conditions. Definition of lateral spread loading conditions requires special geotechnical 
expertise and specific procedures in this area are beyond the scope of this document. 
Defining the loading conditions for fault movement requires specific location of the fault and an 
estimate of the earthquake magnitude on the fault that is consistent with the ground shaking hazard in 
the Provisions. Identification of the fault location should be based on past earthquake movements, 
trenching studies, information from boring logs, or other accepted fault identification techniques. 
Defining fault movement conditions requires special seismological expertise. Additional guidance can 
be found in the Chapter 7 of this Commentary. 
It may not be practically feasible to design a buried structure to resist the effects of permanent ground 
deformation. Alternative approaches in such cases may include relocation to avoid the condition, 
ground improvements to reduce the loads, or implementing special procedures or design features to 
minimize the impact of damage (such as remote controlled or automatic isolation valves that provide the 
ability to rapidly bypass damage or post-earthquake procedures to expedite repair). The goal of 
providing procedures or design features as an alternative to designing for the seismic loadings is to 
change the hazard and function classification of the buried structure such that it is not classified as 
Seismic Use Group II or III. 
It is recommended that one refer to Chapter 7 of this Commentary for additional guidance in 
determining liquefaction potential and determining seismic magnitude. 
Buried structures are subgrade structures such as tanks, tunnels, and pipes. Buried structures that are 
designated as Seismic Use Group II or III, or are of such a size or length to warrant special seismic 
design as determined by the registered design professional, must be identified in the geotechnical report. 
2 003 Commentary, Appendix to Chapter 14 
308 
Buried structures must be designed to resist minimum seismic lateral forces determined from a 
substantiated analysis using approved procedures. Flexible couplings must be provided for buried 
structures requiring special seismic considerations where changes in the support system, configuration, 
or soil condition occur. 
The requirement for and value of flexible couplings should be determined by the properly substantiated 
analysis and approved procedures. It is assumed that the need for flexible couplings refers to buried 
piping or conduits. The prior wording of Sec. A14.2.3 was far too broad in requiring flexible couplings 
where changes in the support system, configuration or soil condition occur. These broad requirements 
could result in flexible couplings installed at locations where permanent ground displacement is 
expected or at transitions between aboveground supported pipe and buried pipe. As currently available 
flexible couplings are not generally designed to match the ultimate strength properties of the piping or 
conduit, the prior requirements potentially introduce a weak point in the piping or conduit system. The 
original focus of the prior requirements was penetrations of buried service lines into a building or other 
structure. Properly designed flexible couplings can be an effective means to limit forces at connections 
to buried structures. However, special care is needed to make sure the design loads and displacements 
are adequately specified. There are several other alternative to providing sufficient flexibility at 
connections to buried structures that are more robust in terms of margin above their design levels. 

 309 
Chapter 15 Commentary 
STRUCTURES WITH DAMPING SYSTEMS 
Background. Chapter 15, Structures with Damping Systems, appears for the first time in the body of 
to the 2003 Provisions, having first appeared as an appendix (to Chapter 13) in the 2000 Provisions. 
The appendix was developed by Technical Subcommittee 12 (TS 12) of the Provision Update 
Committee (PUC) during the 2000 update cycle to provide a basis for designing structures with 
damping systems that is consistent with the NEHRP Provisions, in particular structures with seismic 
(base) isolation systems. Voting members of TS 12 during the 2000 update were Dr. Charles Kircher 
(TS 12 Chair and PUC representative), Dr. Michael Constantinou (PUC representative), Dr. Ian Aiken, 
Dr. Robert Hanson, Mr. Martin Johnson, Dr. Andrew Taylor, and Dr. Andrew Whittaker 
During the 2000 update cycle, the primary resource documents for the design of structures with dampers 
were the NEHRP Guidelines for Seismic Rehabilitation of Buildings (FEMA 273, 1997) and the NEHRP 
Commentary on the Guidelines for the Seismic Rehabilitation of Buildings (FEMA 274, 1997). While 
suitable for the performance-based design, terms, methods of analysis and response limits of the 
NEHRP Guidelines for existing buildings do not match those of the NEHRP Provisions for new 
structures. Accordingly, TS-12 developed new provisions, in particular new linear analysis methods, 
for design of structures with dampers. 
New analysis methods were developed for structures with dampers based on nonlinear pushover 
characterization of the structure and calculation of peak response using effective (secant) stiffness and 
effective damping properties of the first (pushover) mode in the direction of interest. These are same 
concepts used in Chapter 13 to characterize the force-deflection properties of isolation systems, 
modified to explicitly incorporate the effects of ductility demand (post-yield response) and higher-mode 
response of structures with dampers. In contrast to isolated structures, structures with dampers are in 
general expected to yield during strong ground shaking (similar to conventional structures), and their 
performance can be significantly influenced by response of higher modes. 
During the 2000 cycle, analysis methods were evaluated using design examples. Response calculated 
using linear analysis was found to compare well with the results of nonlinear time history analysis 
(Ramirez, 2001). Additional design examples illustrating explicit pushover modeling of the structure 
may be found in Chapter 9 commentary of FEMA 274. The reader is also referred to Ramirez et al. 
(2002a, 2002b, 2003) and Whittaker et al. (2003) for a detailed exposition of the analysis procedures in 
this chapter, background research studies, examples of application and an evaluation of accuracy of the 
linear static and response spectrum analysis methods. 
The balance of this section provides background on the underlying philosophy used by TS12 to develop 
the chapter, the definition the damping system, the concept of effective damping, and the calculation of 
earthquake response using either linear or nonlinear analysis methods. 
Design Philosophy. The basic approach taken by TS12 in developing the chapter for structures with 
damping systems is based on the following concepts: 
1. The chapter is applicable to all types of damping systems, including both displacement-dependent 
damping devices of hysteretic or friction systems and velocity-dependent damping devices of 
viscous or visco elastic systems (Constantinou et al. 1998, Hanson and Soong, 2001) 
2. The chapter provides minimum design criteria with performance objectives comparable to those for 
a structure with a conventional seismic-force-resisting system (but also permits design criteria that 
will achieve higher performance levels).
2 003 Commentary, Chapter 15 
310 
3. The chapter requires structures with a damping system to have a seismic-force-resisting system that 
provides a complete load path. The seismic-force-resisting system must comply with the 
requirements of the Provisions, except that the damping system may be used to meet drift limits. 
4. The chapter requires design of damping devices and prototype testing of damper units for 
displacements, velocities, and forces corresponding to those of the maximum considered earthquake 
(same approach as that used for structures with an isolation system). 
5. The chapter provides linear static and response spectrum analysis methods for design of most 
structures that meet certain configuration and other limiting criteria (for example, at least two 
damping devices at each story configured to resist torsion). The chapter requires additional 
nonlinear response history analysis to confirm peak response for structures not meeting the criteria 
for linear analysis (and for structures close to major faults). 
Damping system. The chapter defines the damping system as: 
The collection of structural elements that includes all individual damping devices, all structural 
elements or bracing required to transfer forces from damping devices to the base of the 
structure, and all structural elements required to transfer forces from damping devices to the 
seismic-force-resisting system. 
The damping system is defined separately from the seismic-force-resisting system, although the two 
systems may have common elements. As illustrated in Figure C15-1, the damping system may be 
external or internal to the structure and may have no shared elements, some shared elements, or all 
elements in common with the seismic-force-resisting system. Elements common to the damping system 
and the seismic-force-resisting system must be designed for a combination of the two loads of the two 
systems. 
The seismic-force-resisting system may be thought of as a collection of lateral-force-resisting elements 
of the structure if the damping system was not functional (as if damping devices were disconnected). 
This system is required to be designed for not less than 75 percent of the base shear of a conventional 
structure (not less than 100 percent, if the structure is highly irregular), using an R factor as defined in 
Table 4.3-1. This system provides both a safety net against damping system malfunction as well as the 
stiffness and strength necessary for the balanced lateral displacement of the damped structure. 
The chapter requires the damping system to be designed for the actual (non-reduced) earthquake forces 
(such as, peak force occurring in damping devices). For certain elements of the damping system, other 
than damping devices, limited yielding is permitted provided such behavior does not affect damping 
system function or exceed the amount permitted by the Provisions for elements of conventional 
structures. 
The chapter defines a damping device as: 
A flexible structural element of the damping system that dissipates energy due to relative 
motion of each end of the device. Damping devices include all pins, bolts, gusset plates, brace 
extensions, and other components required to connect damping devices to other elements of the 
structure. Damping devices may be classified as either displacement-dependent or velocitydependent, 
or a combination thereof, and may be configured to act in either a linear or nonlinear 
manner. 
Following the same approach as that used for design of seismic isolators, damping devices must be 
designed for maximum considered earthquake displacements, velocities, and forces. Likewise, 
prototype damper units must be fully tested to demonstrate adequacy for maximum considered 
earthquake loads and to establish design properties (such as effective damping). 
Structures with Damping Systems 
311 
Internal Damping Devices - Common Elements 
D amper 
SFRS 
DS 
Damper 
SFRS 
DS 
Internal Damping Devices - Some Shared Elements 
Damper 
DS 
SFRS 
D amper 
SFRS 
DS 
External Damping Devices 
Internal Damping Devices - No Shared Elements 
Figure C15-1. Damping system (DS) and seismic-force-resisting system 
(SFRS) configurations. 
Effective Damping. The chapter reduces the response of a structure with a damping system by the 
damping coefficient, B, based on the effective damping, , of the mode of interest. This is the same 
approach as that used by the Provisions for isolated structures. Values of the B coefficient 
recommended for design of damped structures are the same as those in the Provisions for isolated 
structures at damping levels up to 30 percent, but now extend to higher damping levels based on the 
results presented in Ramirez et al. (2001). Like isolation, effective damping of the fundamental-mode 
of a damped structure is based on the nonlinear force-deflection properties of the structure. For use with 
linear analysis methods, nonlinear properties of the structure are inferred from overstrength, O0
, and 
other terms of the Provisions. For nonlinear analysis methods, properties of the structure would be 
based on explicit modeling of the post-yield behavior of elements. 
Figure C15-2 illustrates reduction in design earthquake response of the fundamental mode due to 
effective damping coefficient, B1D. The capacity curve is a plot of the nonlinear behavior of the 
2 003 Commentary, Chapter 15 
312 
fundamental mode in spectral acceleration/displacement coordinates. Damping reduction is applied at 
the effective period of the fundamental mode of vibration (based on the secant stiffness). 
Figure C15-2. Effective damping reduction of design demand. 
In general, effective damping is a combination of three components: 
1. Inherent Damping I
Inherent damping of structure at or just below yield, excluding added viscous 
damping (typically assumed to be 5 percent of critical for structural systems without dampers). 
2. Hysteretic Damping HPost-yield hysteretic damping of the seismic-force-resisting system at the 
amplitude of interest (taken as 0 percent of critical at or below yield). 
3. Added Viscous Damping VViscous component of the damping system (taken as 0 percent for 
hysteretic or friction-based damping systems). 
Both hysteretic damping and the effects of added viscous damping are amplitude-dependent and the 
relative contributions to total effective damping changes with the amount of post-yield response of the 
structure. For example, adding dampers to a structure decreases post-yield displacement of the structure 
and hence decreases the amount of hysteretic damping provided by the seismic-force-resisting system. 
If the displacements were reduced to the point of yield, the hysteretic component of effective damping 
would be zero and the effective damping would be equal to inherent damping plus added viscous 
damping. If there were no damping system (as in a conventional structure), then effective damping 
would simply be equal to inherent damping (typically assumed to be 5 percent of critical for most 
conventional structures). 
Linear Analysis Methods. The chapter specifies design earthquake displacements, velocities, and 
forces in terms of design earthquake spectral acceleration and modal properties. For equivalent lateral 
force (ELF) analysis, response is defined by two modes: (1) the fundamental mode, and (2) the residual 
Spectral Displacement 
Spectral Acceleration 
Capacity 
Curve 
Added Viscous 
plus Inherent 
Damping 
B1D 
Effective 
Damping 
(Added Viscous 
plus Inherent and 
Hysteretic) 
T1D 
T1 
BV+I 
SD1D SDE 
Design Earthquake 
Demand Spectrum 
(V + I Damping) 
Design Earthquake 
Demand Spectrum 
1 (5% Damping) 
D1 
T 
S
1D 
D1 
T
S 
1D 1D 
D1 
T B 
S 
1D 
2 
DS 1D 
2 
1D
D1 1D 
D1 2 B
S T 
4
g 
B
S T 
4
g SD ..
. 
..
. = ..
. 
..
= . 
p p
Structures with Damping Systems 
313 
mode. The residual mode is a new concept used to approximate the combined effects of higher modes. 
While typically of secondary importance to story drift, higher modes can be a significant contributor to 
story velocity and hence are important for design of velocity-dependent damping devices. For response 
spectrum analysis, higher modes are explicitly evaluated. 
For both the ELF and the response spectrum analysis procedures, response in the fundamental mode in 
the direction of interest is based on assumed nonlinear (pushover) properties of the structure. Nonlinear 
(pushover) properties, expressed in terms of base shear and roof displacement, are related to building 
capacity, expressed in terms of spectral coordinates, using mass participation and other fundamentalmode 
factors shown in Figure C15-3. The conversion concepts and factors shown in Figure C15-3 are 
the same as those defined in Chapter 9 of NEHRP Guidelines for the Seismic Rehabilitation of Buildings 
(FEMA 273), which addresses seismic rehabilitation of a structure with damping devices. 
Figure C15-3. Pushover and capacity curves. 
When using linear analysis methods, the shape of the fundamental-mode pushover curve is not known 
and an idealized elasto-plastic shape is assumed, as shown in Figure C15-4. The idealized pushover 
curve shares a common point with the actual pushover curve at the design earthquake displacement, 
D1D. The idealized curve permits defining global ductility demand due to the design earthquake, D, as 
the ratio of design displacement, D1D, to the yield displacement, DY. This ductility factor is used to 
calculate various design factors and to set limits on the building ductility demand, max, which limits are 
consistent with conventional building response limits. Design examples using linear analysis methods 
have been developed and found to compare well with the results of nonlinear time history analysis 
(Ramirez et al., 2001). 
Displacement (in.) 
Acceleration (g) 
Pushover Curve 
Normalized by Weight 
Capacity 
Curve 
( D D) SD ,SA1 1 
..
. 
..
. 
W
D , F D 
1 
1 
S
S
= 
= 
.. 
. 
.. 
. 
= n 
i 
i ii 
n 
i 
i i 
wf 
wf 
W 
1 
2
1 
2 
1 
1 
1 S= 
= n 
i 
i wf 
G W
i 
1 
1 
1 
1 
1 1 1 D SD G D D = 
W
SA W 
W
F 
D 
1 
1 
1 = 
D1 T
2 
1 2 1 1 4 D D D SA T 
p
g SD ..
. 
..
= . 
1 T
2 003 Commentary, Chapter 15 
314 
Figure C15-4. Idealized elasto-plastic pushover curve used for linear analysis. 
The chapter requires elements of the damping system to be designed for actual fundamental-mode 
design earthquake forces corresponding to a base shear value of VY (except that damping devices are 
designed and prototypes tested for maximum considered earthquake response). Elements of the 
seismic-force-resisting system are designed for reduced fundamental-mode base shear, V1, where force 
reduction is based on system overstrength, O0
, conservatively decreased by the ratio, Cd/R, for elastic 
analysis (when actual pushover strength is not known). 
Nonlinear analysis methods. The chapter specifies procedures for the nonlinear response history 
analyses and a nonlinear static procedure. For designs in which the seismic-force-resisting-system will 
remain elastic, only the nonlinear damping device characteristics need to be modeled for these analyses. 
REFERENCES 
Constantinou, M.C., T.T. Soong and G.F. Dargush. 1998. Passive Energy Dissipation Systems for 
Structural Design and Retrofit, Monograph No. 1, Multidisciplinary Center for Earthquake Engineering 
Research, University of Buffalo, State University of New York, Buffalo, NY. 
Hanson, Robert D. and Tsu T. Soong. 2001. Seismic Design with Supplemental Energy Dissipation 
Devices, MNO-8, Earthquake Engineering Research Institute, Oakland, CA. 
Ramirez, O.M., M.C. Constantinou, C.A. Kircher, A. Whittaker, M. Johnson, J.D. Gomez and C.Z. 
Chrysostomou. 2001. Development and Evaluation of Simplified Procedures of Analysis and Design for 
Structures with Passive Energy Dissipation Systems, Technical Report MCEER-00-0010, Revision 1, 
Multidisciplinary Center for Earthquake Engineering Research, University of Buffalo, State University 
of New York, Buffalo, NY. 
Roof Displacement 
Base Shear 
DY D1D 
VY 
V1 
Overstrength 
(nominal value, OO, 
reduced by CD/R 
ratio when used 
with elasto-plastic 
curve) 
O0Cd 
R O0 
2 max 
1
2 
1D 
Y 
1D 
D T
T 
D
 = D  = 
1D 
2 
DS 1D 
2 1 
1D
D1 1D 
1D 2 1 B
G S T 
4p
g 
B
G S T 
4p
g D ..
. 
..
. = ..
. 
..
= . 
D 
Pushover 
Curve 
Idealized 
Elasto-Plastic 
Pushover 
Curve
Structures with Damping Systems 
315 
Ramirez, O.M., Constantinou, M.C., Gomez, J., Whittaker, A.S., and Chrysostomou, C.Z. 2002a. 
Evaluation of Simplified Methods of Analysis of Yielding Structures With Damping Systems, Earthquake 
Spectra, Vol. 18, No. 3, Aug., pp. 501-530. 
Ramirez, O.M., Constantinou, M.C., Whittaker, A.S., Kircher, C.A., and Chrysostomou, C.Z. 2002b. 
Elastic And Inelastic Seismic Response of Buildings With Damping Systems, Earthquake Spectra, Vol. 
18, No. 3, Aug., pp. 531-547. 
Ramirez, O.M., Constantinou, M.C., Whittaker, A.S., Kircher, C.A., Johnson, M.W. and 
Chrysostomou, C.Z. 2003. Validation Of 2000 NEHRP Provisions Equivalent Lateral Force and 
Modal Analysis Procedures For Buildings With Damping Systems, Earthquake Spectra, Vol. 19, No. 4, 
November, pp. 981-999. 
Whittaker, A.S., Constantinou, M.C., Ramirez, O.M., Johnson, M.W. and Chrysostomou, C.Z. 2003. 
Equivalent Lateral Force and Modal Analysis Procedures of the 2000 NEHRP Provisions For Buildings 
with Damping Systems, Earthquake Spectra, Vol. 19, No. 4, November, pp. 959-980. 
2 003 Commentary, Chapter 15 
316 
This page is intentionally left blank. 

 317 
Commentary Appendix A 
DEVELOPMENT OF MAXIMUM CONSIDERED EARTHQUAKE 
GROUND MOTION MAPS FIGURES 3.3-1 THROUGH 3.3-14 
BACKGROUND 
The maps used in the Provisions through 1994 provided the Aa (effective peak acceleration 
coefficient) and Av (effective peak velocity-related acceleration coefficient) values to use for design. 
The BSSC had always recognized that the maps and coefficients would change with time as the 
profession gained more knowledge about earthquakes and their resulting ground motions and as 
society gained greater insight into the process of establishing acceptable risk. 
By 1997, significant additional earthquake data had been obtained that made the Aa and Av maps, then 
about 20 years old, seriously out of date. For the 1997 Provisions, a joint effort involving the BSSC, 
the Federal Emergency Management Agency (FEMA), and the U.S. Geological Survey (USGS) was 
conducted to develop both new maps for use in design and new design procedures reflecting the 
significant advances made in the past 20 years. The BSSCs role in this joint effort was to develop 
new ground motion maps for use in design and design procedures based on new USGS seismic hazard 
maps. 
The BSSC appointed a 15-member Seismic Design Procedure Group (SDPG) to develop the seismic 
ground motion maps and design procedures. The SDPG membership was composed of representatives 
of different segments of the design community as well as two earth science members designated by the 
USGS, and the membership was representative of the different geographical regions of the country. 
Also, the BSSC, with input from FEMA and USGS, appointed a five-member Management 
Committee (MC) to guide the efforts of the SDPG. The MC was geographically balanced insofar as 
practicable and was composed of two seismic hazard definition experts and three engineering design 
experts, including the chairman of the 1997 Provisions Update Committee (PUC). The SDPG and the 
MC worked closely with the USGS to define the BSSC mapping needs and to understand how the 
USGS seismic hazard maps should be used to develop the BSSC seismic ground motion maps and 
design procedures. 
For a brief overview of how the USGS developed its hazard maps, see Appendix B to this 
Commentary volume. A detailed description of the development of the maps is contained in the 
USGS Open-File Report 96-532, National Seismic-Hazard Maps: Documentation, June 1996, by 
Frankel, et al. (1996). The USGS hazard maps also can be viewed and printed from a USGS Internet 
site at http://eqhazmaps.usgs.gov. 
The goals of the SDPG were as follows: 
1. To replace the existing effective peak acceleration and velocity-related acceleration design maps 
with new ground motion spectral response maps based on new USGS seismic hazard maps. 
2. To develop the new ground motion spectral response maps within the existing framework of the 
Provisions with emphasis on uniform margin against the collapse of structures. 
3. To develop design procedures for use with the new ground motion spectral response maps. 
PURPOSE OF THE PROVISIONS 
The purpose of the Provisions is to present criteria for the design and construction of new structures 
subject to earthquake ground motions in order to minimize the risk to life for all structures, to increase 
2003 Commentary, Appendix A 
318 
the expected performance of higher occupancy structures as compared to ordinary structures, and to 
improve the capability of essential structures to function after an earthquake. To this end, the 
Provisions provide the minimum criteria considered prudent for structures subjected to earthquakes at 
any location in the United States and its territories. The Provisions generally considers property 
damage as it relates to occupant safety for ordinary structures. For high occupancy and essential 
structures, damage limitation criteria are more strict in order to better provide for the safety of 
occupants and the continued functioning of the structure. Some structural and nonstructural damage 
can be expected as a result of the design ground motions because the Provisions allow inelastic 
energy dissipation by utilizing the deformability of the structural system. For ground motions in 
excess of the design levels, the intent is that there be a low likelihood of collapse. These goals of the 
Provisions were the guiding principles for developing the design maps. 
POLICY DECISIONS FOR SEISMIC GROUND MOTION MAPS 
The new maps (cited in both the 1997 and 2000 Provisions) reflect the following policy decisions that 
depart from past practice and the 1994 Provisions: 
1. The maps define the maximum considered earthquake ground motion for use in design procedures, 
2. The use of the maps for design provide an approximately uniform margin against collapse for 
ground motions in excess of the design levels in all areas. 
3. The maps are based on both probabilistic and deterministic seismic hazard maps, and 
4. The maps are response spectra ordinate maps and reflect the differences in the short-period range 
of the response spectra for the areas of the United States and its territories with different ground 
motion attenuation characteristics and different recurrence times. 
These policy decisions reflected new information from both the seismic hazard and seismic 
engineering communities that is discussed below. 
In the 1994 Provisions, the design ground motions were based on an estimated 90 percent probability 
of not being exceeded in 50 years (about a 500 year mean recurrence interval) (ATC 3-06 1978). The 
1994 Provisions also recognized that larger ground motions are possible and that the larger motions, 
although their probability of occurrence during a structures life is very small, nevertheless can occur 
at any time. The 1994 Provisions also defined a maximum capable earthquake as the maximum 
level of earthquake ground shaking that may ever be expected at the building site within the known 
geologic framework. It was additionally specified that in certain map areas ($ Aa = 0.3), the 
maximum capable earthquake was associated with a motion that has a 90 percent probability of not 
being exceeded in 100 years (about a 1000 year mean recurrence interval). In addition to the 
maximum capable earthquake definition, sample ground motion maps were prepared with 90 percent 
probabilities of not being exceeded in 250 years (about a 2500 year mean recurrence interval). 
Given the wide range in return periods for maximum magnitude earthquakes throughout the United 
States and its territories (100 years in parts of California to 100,000 years or more in several other 
locations), current efforts have focused on defining the maximum considered earthquake ground 
motions for use in design (not the same as the maximum capable earthquake defined in the 1994 
Provisions). The maximum considered earthquake ground motions are determined in a somewhat 
different manner depending on the seismicity of an individual region; however, they are uniformly 
defined as the maximum level of earthquake ground shaking that is considered as reasonable to design 
structures to resist. Focusing on ground motion versus earthquake size facilitates the development of a 
design approach that provides an approximately uniform margin against collapse throughout the 
United States. 
As noted above, the 1994 Provisions generally used the notation of 90 percent probability of not being 
exceeded in a certain exposure time period (50, 100, or 250 years), which can then be used to calculate 
Development of Maximum Considered Earthquake Ground Motion Maps Figures 3.3-1 through 3.314 
319 
a given mean recurrence interval (500, 1000, or 2500 years). For the purpose of the new maps and 
design procedure introduced in the 1997 Provisions, the single exposure time period of 50 years has 
been commonly used as a reference period over which to consider loads on structures (after 50 years 
of use, structures may require evaluation to determine future use and rehabilitation needs). With this 
in mind, different levels of probability or return period are expressed as percent probability of 
exceedance in 50 years. Specifically, 10 percent probability of exceedance in 50 years is a mean 
recurrence interval of about 500 years, 5 percent probability of exceedance in 50 years is a mean 
recurrence interval of about 1000 years, and 2 percent probability of exceedance in 50 years is a mean 
recurrence interval of about 2500 years. The above notation is used throughout the Provisions. 
Review of modern probabilistic seismic hazard results, including the maps prepared by the USGS to 
support the effort resulting in the 1997 Provisions, indicates that the rate of change of ground motion 
versus probability is not constant throughout the United States. For example, the ground motion 
difference between the 10 percent probability of exceedance and 2 percent probability of exceedance 
in 50 years in coastal California is typically smaller than the difference between the two probabilities 
in less active seismic areas such as the eastern or central United States. Because of these differences, 
questions were raised concerning whether definition of the ground motion based on a constant 
probability for the entire United States would result in similar levels of seismic safety for all 
structures. Figure A1 plots the 0.2 second spectral acceleration normalized at 2 percent probability of 
exceedance in 50 years versus the annual frequency of exceedance. Figure A1 shows that in coastal 
California, the ratio between the 0.2 second spectral acceleration for the 2 and the 10 percent 
probabilities of exceedance in 50 years is about 1.5 whereas, in other parts of the United States, the 
ratio varies from 2.0 to 5.0. 
FIGURE A1 Relative hazard at selected sites for 0.2 sec spectral response acceleration. The hazard 
curves are normalized at 2 percent probability of exceedance in 50 years. 

2003 Commentary, Appendix A 
320 
In answering the questions, it was recognized that seismic safety is the result of a number of steps in 
addition to defining the design earthquake ground motions, including the critical items generally 
defined as proper site selection, structural design criteria, analysis and procedures, detailed design 
requirements, and construction. The conservatism in the actual design of the structure is often referred 
to as the seismic margin. It is the seismic margin that provides confidence that significant loss of 
life will not be caused by actual ground motions equal to the design levels. Alternatively, the seismic 
margin provides a level of protection against larger, less probable earthquakes although at a lower 
level of confidence. 
The collective opinion of the SDPG was that the seismic margin contained in the Provisions provides, 
as a minimum, a margin of about 1.5 times the design earthquake ground motions. In other words, if a 
structure experiences a level of ground motion 1.5 times the design level, the structure should have a 
low likelihood of collapse. The SDPG recognizes that quantification of this margin is dependent on 
the type of structure, detailing requirements, etc., but the 1.5 factor is a conservative judgment 
appropriate for structures designed in accordance with the Provisions. This seismic margin estimate is 
supported by Kennedy et al. (1994), Cornell (1994), and Ellingwood (1994) who evaluated structural 
design margins and reached similar conclusions. 
The USGS seismic hazard maps indicate that in most locations in the United States the 2 percent 
probability of exceedance in 50 years ground motion values are more than 1.5 times the 10 percent 
probability of exceedance in 50 years ground motion values. This means that if the 10 percent 
probability of exceedance in 50 years map was used as the design map and the 2 percent probability of 
exceedance in 50 years ground motions were to occur, there would be low confidence (particularly in 
the central and eastern United States) that structures would not collapse due to these larger ground 
motions. Such a conclusion for most of the United States was not acceptable to the SDPG. The only 
location where the above results seemed to be acceptable was coastal California (2 percent probability 
of exceedance in 50 years map is about 1.5 times the 10 percent probability of exceedance in 50 years 
map) where structures have experienced levels of ground shaking equal to and above the design value. 
The USGS probabilistic seismic hazard maps for coastal California also indicate the 10 percent 
probability of exceedance in 50 years seismic hazard map is significantly different from (in most cases 
larger) the design ground motion values contained in the 1994 Provisions. Given the generally 
successful experience with structures that complied with the recent editions of the Uniform Building 
Code whose design map contained many similarities to the 1994 Provisions design map, the SDPG 
was reluctant to suggest large changes without first understanding the basis for the changes. This 
stimulated a detailed review of the probabilistic maps for coastal California. This review identified a 
unique issue for coastal California in that the recurrence interval of the estimated maximum magnitude 
earthquake is less than the recurrence interval represented on the probabilistic map, in this case the 10 
percent probability of exceedance in 50 years map (i.e., recurrence interval for maximum magnitude 
earthquake is 100 to 200 years versus 500 years.) 
Given the above, one choice was to accept the change and use the 10 percent probability of 
exceedance in 50 years probabilistic map to define the design ground motion for coastal California 
and, using this, determine the appropriate probability for design ground motion for the rest of the 
United States that would result in the same level of seismic safety. This would have resulted in the 
design earthquake being defined at 2 percent probability of exceedance in 50 years and the need for 
development of a 0.5 to 1.0 percent probability of exceedance in 50 years map to show the potential 
for larger ground motions outside of coastal California. Two major problems were identified. The 
first is that requiring such a radical change in design ground motion in coastal California seems to 
contradict the general conclusion that the seismic design codes and process are providing an adequate 
level of life safety. The second is that completing probabilistic estimates of ground motion for lower 
probabilities (approaching those used for critical facilities such as nuclear power plants) is associated 
with large uncertainties and can be quite controversial. 
Development of Maximum Considered Earthquake Ground Motion Maps Figures 3.3-1 through 3.314 
321 
An alternative choice was to build on the observation that the maximum earthquake for many seismic 
faults in coastal California is fairly well known and associated with probabilities larger than a 10 
percent probability of exceedance in 50 years (500 year mean recurrence interval). Given this, a 
decision was made to develop a procedure that would use the best estimate of ground motion from 
maximum magnitude earthquakes on seismic faults with high probabilities of occurrence (short return 
periods). For the purposes of the Provisions, these earthquakes are defined as deterministic 
earthquakes. Following this approach and recognizing the inherent seismic margin contained in the 
Provisions, it was determined that the level of seismic safety achieved in coastal California would be 
approximately equivalent to that associated with a 2 to 5 percent probability of exceedance in 50 years 
for areas outside of coastal California. In other words, the use of the deterministic earthquakes to 
establish the maximum considered earthquake ground motions for use in design in coastal California 
results in a level of protection close to that implied in the 1994 Provisions and consistent with 
maximum magnitude earthquakes expected for those seismic sources. Additionally, this approach 
results in less drastic changes to ground motion values for coastal California than the alternative 
approach of using probabilistic based maps. 
One could ask why any changes are necessary for coastal California given the positive experience 
from recent earthquakes. While it is true that the current seismic design practices have produced 
positive results, the current design ground motions in the 1994 Provisions are less than those expected 
from maximum magnitude earthquakes on known seismic sources. The 1994 Provisions reportedly 
considered maximum magnitude earthquakes but did not directly link them to the design ground 
motions (Applied Technology Council, 1978). If there is high confidence in the definition of the fault 
and magnitude of the earthquake and the maximum earthquake occurs frequently, then the design 
should be linked to at least the best estimate ground motion for such an earthquake. Indeed, it is the 
actual earthquake experience in coastal California that is providing increased confidence in the seismic 
margins contained in the Provisions. 
The above approach also is responsive to comments that the use of 10 percent probability of 
exceedance in 50 years is not sufficiently conservative in the central and eastern United States where 
the earthquakes are expected to occur infrequently. Based on the above discussion and the inherent 
seismic margin contained in the Provisions, the SDPG selected 2 percent probability of exceedance in 
50 years as the maximum considered earthquake ground motion for use in design where the use of the 
deterministic earthquake approach discussed above is not used. 
The maximum considered earthquake ground motion maps are based on two response spectral values 
(a short-period and a long-period value) instead of the Aa and Av coefficients. The decision to use 
response spectral values is based on earthquake data obtained during the past 20 years showing that 
site-specific spectral values are more appropriate for design input than the Aa and Av coefficients used 
with standardized spectral shapes. The spectral shapes vary in different areas of the country and for 
different site conditions. This is particularly the case for the short-period portion of the response 
spectra. Based on the differences in the ground motion attenuation characteristics between the central 
and eastern and western United States, the USGS used different ground motion attenuation functions 
for these areas in developing the seismic hazard maps. The ground motion attenuation functions in the 
eastern United States result in higher short-period spectral accelerations at lower periods for a given 
earthquake magnitude than the western United States attenuation functions, particularly compared to 
the high seismicity region of coastal California. The short-period response spectral values were 
reviewed in order to determine the most appropriate value to use for the maximum considered 
earthquake ground motion maps. Based on this review, the short-period spectral response value of 0.2 
second was selected to represent the short-period range of the response spectra for the eastern United 
States. In the western United States the most appropriate short-period response spectral value was 
determined to be 0.3 second, but a comparison of the 0.2 and 0.3 second values indicated that the 
differences in the response spectral values were insignificant. Based on this and for convenience of 
preparing the maximum considered earthquake ground motion maps, the short-period response 
spectral value of 0.2 second was selected to represent the short-period range of the response spectra 
2003 Commentary, Appendix A 
322 
for all of the United States. The long-period response spectral value selected for use is 1.0 second for 
all of the United States. Based on the ground motion attenuation functions and the USGS seismic 
hazard maps, a 1/T (T = natural period) relationship was selected to define the response spectra from 
the short period value to the long-period value. Using the spectral values from the ground motion 
maps will allow the different spectral shapes to be incorporated into design. 
DEVELOPMENT OF THE MAXIMUM CONSIDERED EARTHQUAKE GROUND MOTION 
MAPS FOR USE IN DESIGN 
The concept for developing maximum considered earthquake ground motions for use in design 
involved two distinct steps: 
1. The various USGS probabilistic seismic hazard maps were combined with deterministic hazard 
maps by a set of rules (logic) to create the maximum considered earthquake ground motion maps 
that can be used to define response spectra for use in design and 
2. Design procedures were developed that transform the response spectra into design values (e.g., 
design base shear). 
The response spectra defined from the first step represent general site-dependent spectra similar to 
those that would be obtained by a geotechnical study and used for dynamic analysis except their 
shapes are less refined (i.e., shape defined for only a limited number of response periods). The 
response spectra do not represent the same hazard level across the country but do represent actual 
ground motion consistent with providing approximately uniform protection against the collapse of 
structures. The response spectra represent the maximum considered earthquake ground motions for 
use in design for Site Class B (rock with a shear wave velocity of 760 meters/second). 
The maximum considered earthquake ground motion maps for use in design are based on a defined set 
of rules for combining the USGS seismic hazard maps to reflect the differences in the ability to define 
the fault sources and seismicity characteristics across the regions of the country as discussed in the 
policy decisions. Accommodating regional differences allows the maximum considered earthquake 
maps to represent ground motions for use in design that provide reasonably consistent margins of 
preventing the collapse of structures. Based on this, three regions have been defined: 
1. Regions of negligible seismicity with very low probability of collapse of the structure, 
2. Regions of low and moderate to high seismicity, and 
3. Regions of high seismicity near known fault sources with short return periods. 
Regions of Negligible Seismicity With Very Low Probability of Collapse of the Structure 
The regions of negligible seismicity with very low probability of collapse have been defined by: 
1. Determining areas where the seismic hazard is controlled by earthquakes with Mb (body wave 
magnitude) magnitudes less than or equal to 5.5 and 
2. Examining the recorded ground motions associated with Modified Mercalli Intensity V. 
The basis for the first premise is that in this region, there are a number of examples of earthquakes 
with Mb  5.5 which caused only localized damage to structures not designed for earthquakes. The 
basis for the second premise is that Modified Mecalli Intensity V ground motions typically do not 
cause structural damage. By definition, Modified Mercalli Intensity V ground shaking is felt by most 
people, displaces or upsets small objects, etc., but typically causes no, or only minor, structural 
damage in buildings of any type. Modified Mercalli Intensity VI ground shaking is felt by everyone, 
small objects fall off shelves, etc., and minor or moderate structural damage occurs to weak plaster and 
masonry construction. Life-threatening damage or collapse of structures would not be expected for 
either Modified Mercalli Intensities V or VI ground shaking. Based on an evaluation of 1994 
Northridge earthquake data, regions of different Modified Mercalli Intensity (Dewey, 1995) were 
correlated with maps of smooth response spectra developed from instrumental recordings 
Development of Maximum Considered Earthquake Ground Motion Maps Figures 3.3-1 through 3.314 
323 
(Sommerville, 1995). The Northridge earthquake provided a sufficient number of instrumental 
recordings and associated spectra to permit correlating Modified Mercalli Intensity with response 
spectra. The results of the correlation determined the average response spectrum for each Modified 
Mercalli Intensity region. For Modified Mercalli Intensity V, the average response spectrum of that 
region had a spectral response acceleration of slightly greater than 0.25g at 0.3 seconds and a spectral 
response acceleration of slightly greater than 0.10g at 1.0 seconds. On the basis of these values and 
the minor nature of damage associated with Modified Mercalli Intensity V, 0.25g (short-period 
acceleration) and 0.10g (acceleration at a period of 1 second, taken proportional to 1/T) is deemed to 
be a conservative estimate of the spectrum below which life-threatening damage would not be 
expected to occur even to the most vulnerable of types of structures. Therefore, this region is defined 
as areas having maximum considered earthquake ground motions with a 2 percent probability of 
exceedance in 50 years equal to or less than 0.25g (short period) and 0.10g (long period). The seismic 
hazard in these areas is generally the result of Mb  5.5 earthquakes. In these areas, a minimum lateral 
force design of 1 percent of the dead load of the structure shall be used in addition to the detailing 
requirements for the Seismic Design Category A structures. 
In these areas it is not considered necessary to specify seismic-resistant design on the basis of a 
maximum considered earthquake ground motion. The ground motion computed for such areas is 
determined more by the rarity of the event with respect to the chosen level of probability than by the 
level of motion that would occur if a small but close earthquake actually did occur. However, it is 
desirable to provide some protection, both against earthquakes as well as many other types of 
unanticipated loadings. The requirements for Seismic Design Category A provide a nominal amount 
of structural integrity that will improve the performance of buildings in the event of a possible, but 
rare earthquake. The result of design to Seismic Design Category A is that fewer buildings would 
collapse in the vicinity of such an earthquake. 
The integrity is provided by a combination of requirements. First, a complete load path for lateral 
forces must be identified. Then it must be designed for a lateral force equal to a 1% acceleration on 
the mass. Lastly, the minimum connection forces specified for Seismic Design Category A must be 
satisfied. 
The 1 percent value has been used in other countries as a minimum value for structural integrity. For 
many structures, design for the wind loadings specified in the local building codes will normally 
control the lateral force design when compared to the minimum structural integrity force on the 
structure. However, many low-rise heavy structures or structures with significant dead loads resulting 
from heavy equipment may be controlled by the nominal 1 percent acceleration. Also, minimum 
connection forces may exceed structural forces due to wind in additional structures. 
The regions of negligible seismicity will vary depending on the Site Class on which structures are 
located. The Provisions seismic ground motion maps (Maps 1 through 19 ) are for Site Class B 
conditions and the region of negligible seismicity for Site Class B is defined where the maximum 
considered earthquake ground motion short-period values are #0.25g and the long-period values are # 
0.10g. The regions of negligible seismicity for the other Site Classes are defined by using the 
appropriate site coefficients to determine the maximum considered earthquake ground motion for the 
Site Class and then determining if the short-period values are # 0.25g and the long-period values are # 
0.10g. If so, then the site of the structure is located in the region of negligible seismicity for that Site 
Class. 
Regions of Low and Moderate to High Seismicity 
In regions of low and moderate to high seismicity, the earthquake sources generally are not well 
defined and the maximum magnitude estimates have relatively long return periods. Based on this, 
probabilistic hazard maps are considered to be the best means to represent the uncertainties and to 
define the response spectra for these regions. The maximum considered earthquake ground motion for 
2003 Commentary, Appendix A 
324 
these regions is defined as the ground motion with a 2 percent probability of exceedance in 50 years. 
The basis for this decision is explained in the policy discussion. 
Consideration was given to establishing a separate region of low seismicity and defining a minimum 
level of ground motion (i.e., deterministic minimum ground motions). This was considered because in 
the transition between the regions of negligible seismicity to the regions of low seismicity, the ground 
motions are relatively small and may not be very meaningful for use in seismic design. The minimum 
level was also considered because the uncertainty in the ground motion levels in the regions of low 
seismicity is larger than in the regions of moderate to high seismicity. This larger uncertainty may 
warrant consideration of using higher ground motions (or some minimum level of ground motion) than 
provided by the maximum considered earthquake ground motions shown on the maps. 
The studies discussed above for the regions of negligible seismicity by Dewey (1995) and 
Sommerville (1995), plus other unpublished studies (to date), were evaluated as a means of 
determining minimum levels of ground motion for used in design. These studies correlated the 
Modified Mercalli Intensity data with the recorded ground motions and associated damage. The 
studies included damage information for a variety of structures which had no specific seismic design 
and determined the levels of ground motion associated with each Modified Mercalli Intensity. These 
studies indicate that ground motion levels of about 0.50g short-period spectral response and 0.20g 
long-period spectral response are representative of Modified Mercalli Intensity VII damage. 
Modified Mercalli Intensity VII ground shaking results in negligible damage in buildings of good 
design and construction, slight to moderate damage in well-built ordinary buildings, considerable 
damage in poorly-built or badly designed buildings, adobe houses, old walls (especially where laid up 
without mortar), etc. In other words, Modified Mercalli VII ground shaking is about the level of 
ground motion where significant structural damage may occur and result in life safety concerns for 
occupants. This tends to suggest that designing structures for ground motion levels below 0.50g shortperiod 
spectral response and 0.20g long-period spectral response may not be meaningful. 
One interpretation of this information suggests that the ground motion levels for defining the regions 
of negligible seismicity could be increased. This interpretation would result in much larger regions 
that require no specific seismic design compared to the 1994 Provisions. 
Another interpretation of the information suggests establishing a minimum level of ground Motion (at 
about the Modified Mercalli VII shaking) for regions of low seismicity, in order to transition from the 
regions of negligible seismicity to the region of moderate to high seismicity. Implementation of a 
minimum level of ground motion, such as 0.50g for the short-period spectral response and 0.20g for 
the long-period spectral response, would result in increases (large percentages) in ground motions used 
for design compared to the 1994 Provisions. 
Based on the significant changes in past practices resulting from implementing either of the above 
interpretations, the SDPG decided that additional studies are needed to support these changes. Results 
of such studies should be considered for future editions of the Provisions. 
Regions of High Seismicity Near Known Fault Sources With Short Return Periods 
In regions of high seismicity near known fault sources with short return periods, deterministic hazard 
maps are used to define the response spectra maps as discussed above. The maximum considered 
earthquake ground motions for use in design are determined from the USGS deterministic hazard 
maps developed using the ground motion attenuation functions based on the median estimate 
increased by 50 percent. Increasing the median ground motion estimates by 50 percent is deemed to 
provide an appropriate margin and is similar to some deterministic estimates for a large magnitude 
characteristic earthquake using ground motion attenuation functions with one standard deviation. 
Estimated standard deviations for some active fault sources have been determined to be higher than 50 
percent, but this increase in the median ground motions was considered reasonable for defining the 
maximum considered earthquake ground motions for use in design. 
Development of Maximum Considered Earthquake Ground Motion Maps Figures 3.3-1 through 3.314 
325 
Maximum Considered Earthquake Ground Motion Maps for Use in Design 
Considering the rules for the three regions discussed above, the maximum considered earthquake 
ground motion maps for use in design were developed by combining the regions in the following 
manner: 
1. Where the maximum considered earthquake map ground motion values (based on the 2 percent 
probability of exceedance in 50 years) for Site Class B adjusted for the specific site conditions are 
# 0.25g for the short-period spectral response and # 0.10g for the long period spectral response, 
then the site will be in the region of negligible seismicity and a minimum lateral force design of 1 
percent of the dead load of the structure shall be used in addition to the detailing requirements for 
the Seismic Design Category A structures. 
2. Where the maximum considered earthquake ground motion values (based on the 2 percent 
probability of exceedance in 50 years) for Site Class B adjusted for the specific site conditions are 
greater than 0.25g for the short-period spectral response and 0.10g for the long-period spectral 
response, the maximum considered earthquake ground motion values (based on the 2 percent 
probability of exceedance in 50 years adjusted for the specific site conditions) will be used until 
the values equal the present (1994 Provisions) ceiling design values increased by 50 percent (short 
period = 1.50g, long period = 0.60g). The present ceiling design values are increased by 50 
percent to represent the maximum considered earthquake ground motion values. This will define 
the sites in regions of low and moderate to high seismicity. 
3. To transition from regions of low and moderate to high seismicity to regions of high seismicity 
with short return periods, the maximum considered earthquake ground motion values based on 2 
percent probability of exceedance in 50 years will be used until the values equal the present (1994 
Provisions) ceiling design values increased by 50 percent (short period = 1.50g, long period = 
0.60g). The present ceiling design values are increased by 50 percent to represent maximum 
considered earthquake ground motion values. When the 1.5 times the ceiling values are reached, 
then they will be used until the deterministic maximum considered earthquake map values of 1.5g 
(long period) and 0.60g (short period) are obtained. From there, the deterministic maximum 
considered earthquake ground motion map values will be used. 
In some cases there are regions of high seismicity near known faults with return periods such that the 
probabilistic map values (2 percent probability of exceedance in 50 years) will exceed the present 
ceiling values of the 1994 Provisions increased by 50 percent and will be less than the deterministic 
map values. In these regions, the probabilistic map values will be used for the maximum considered 
earthquake ground motions. 
The basis for using present ceiling design values as the transition between the two regions is because 
earthquake experience has shown that regularly configured, properly designed structures performed 
satisfactorily in past earthquakes. The most significant structural damage experienced in the 
Northridge and Kobe earthquakes was related to configuration, structural systems, inadequate 
connection detailing, incompatibility of deformations, and design or construction deficiencies -- not 
due to deficiency in strength (Structural Engineers Association of California, 1995). The earthquake 
designs of the structures in the United States (coastal California) which have performed satisfactorily 
in past earthquakes were based on the criteria in the Uniform Building Code. Considering the site 
conditions of the structures and the criteria in the Uniform Building Code, the ceiling design values for 
these structures were determined to be appropriate for use with the Provisions maximum considered 
earthquake ground motion maps for Site Class B. Based on this, the equivalent maximum considered 
earthquake ground motion values for the ceiling were determined to be 1.50g for the short period and 
0.60g for the long period. 
As indicated above there also are some regions of high seismicity near known fault sources with return 
periods such that the probabilistic map values (2 percent probability of exceedance in 50 years) will 
exceed the ceiling values of the 1994 Provisions increased by 50 percent and also be less than the 
2003 Commentary, Appendix A 
326 
deterministic map values. In these regions, the probabilistic map values are used for the maximum 
considered earthquake ground motions. 
The near source area in the high seismicity regions is defined as the area where the maximum 
considered earthquake ground motion values are $ 0.75g on the 1.0 second map. In the near source 
area, Provisions Sec. 5.2.3 through 5.2.6 impose additional requirements for certain structures unless 
the structures are fairly regular, do not exceed 5 stories in height, and do not have a period of vibration 
over 0.5 seconds. For the fairly regular structures not exceeding 5 stories in height and not having a 
period of vibration over 0.5 seconds, the maximum considered earthquake ground motion values will 
not exceed the present ceiling design values increased by 50 percent. The basis for this is because of 
the earthquake experience discussed above. 
These development rules for the maximum considered earthquake ground motion maps for use in 
design are illustrated in Figures A2 and A3. The application of these rules resulted in the maximum 
considered earthquake ground motion maps (Maps 1 through 24) introduced in the 1997 and used 
again in the 2000 Provisions. 
FIGURE A2 Development of the maximum considered earthquake 
ground motion map for spectral acceleration of T = 1.0, Site Class B. 
Development of Maximum Considered Earthquake Ground Motion Maps Figures 3.3-1 through 3.314 
327 
STEP 1 -- DEFINE POTENTIAL SEISMIC SOURCES 
A. Compile Earth Science Information -- Compile historic seismicity and fault characteristics 
including earthquake magnitudes and recurrence intervals. 
B. Prepare Seismic Source Map -- Specify historic seismicity and faults used as sources. 
STEP 2 -- PREPARE PROBABILISTIC AND DETERMINISTIC SPECTRAL RESPONSE 
MAPS 
A. Develop Regional Attenuation Relations 
(1) Eastern U.S. ( Toro, et al., 1993, and Frankel, 1996) 
(2) Western U.S. (Boore et al., 1993 &1994, Campbell and Bozorgnia, 1994, and Sadigh, 1993 
for PGA. Boore et al., 1993 &1994, and Sadigh, 1993 for spectral values) 
(3) Deep Events (35km) (Geomatrix et al., 1993) 
(4) Cascadia Subduction Zone (Geomatrix et al., 1993, and Sadigh, 1993) 
B. Prepare Probabilistic Spectral Response Maps (USGS Probabilistic Maps) -- Maps showing SS 
and S1 where SS and S1 are the short and 1 second period ground motion response spectral 
values for a 2 percent chance of exceedence in 50 years inferred for sites with average shear wave 
velocity of 760 m/s from the information developed in Steps 1A and 1B and the ground motion 
attenuation relationships in Step 2A. 
C. Prepare Deterministic Spectral Response Maps (USGS Deterministic Maps) -- Maps showing SS 
and S1 for faults and maximum earthquakes developed in Steps 1A and 1B and the median ground 
motion attenuation relations in Step 2A increased by 50% to represent the uncertainty. 
STEP 3 -- PREPARE EARTHQUAKE GROUND MOTION SPECTRAL RESPONSE MAPS 
FOR PROVISIONS (MAXIMUM CONSIDERED EARTHQUAKE GROUND MOTION MAP) 
Region 1 -- Regions of Negligible Seismicity with Very Low Probability of Collapse of the 
Structure (No Spectral Values) 
Region definition: Regions for which SS < 0.25g and S1 < 0.10g from Step 2B. 
Design values: No spectral ground motion values required. Use a minimum lateral force level of 1 
percent of the dead load for Seismic Design Category A. 
Region 2 -- Regions of Low and Moderate to High Seismicity (Probabilistic Map Values) 
Region definition: Regions for which 0.25g < SS < 1.5g and 0.25g < S1 < 0.60g from Step 2B. 
Maximum considered earthquake map values: Use SS and SI map values from Step 2B. 
Transition Between Regions 2 and 3 - Use MCE values of SS = 1.5g and S1 = 0.60g 
Region 3 -- Regions of High Seismicity Near Known Faults (Deterministic Values) 
Region definition: Regions for which 1.5g < SS and 0.60g < S1 from Step 2C. 
Maximum considered earthquake map values: Use SS and SI map values from Step 2C. 
FIGURE A3 Methodology for development of the maximum considered earthquake ground 
motion maps (Site Class B). 

2003 Commentary, Appendix A 
328 
Use of the Maximum Considered Earthquake Ground Motion Maps in the Design Procedure: 
The 1994 Provisions defined the seismic base shear as a function of the outdated effective peak 
velocity-related acceleration Av, and effective peak acceleration, Aa. Beginning with the 1997 
Provisions, the base shear of the structure is defined as a function of the maximum considered 
earthquake ground motion maps where SS = maximum considered earthquake spectral acceleration in 
the short-period range for Site Class B; S1 = maximum considered earthquake spectral acceleration at 
the 1.0 second period for Site Class B; SMS = FaSS, maximum considered earthquake spectral 
acceleration in the short-period range adjusted for Site Class effects where Fa is the site coefficient 
defined in Provisions Sec. 4.1.2; SM1 = FvS1, maximum considered earthquake spectral acceleration at 
1.0 second period adjusted for Site Class effects where Fv is the site coefficient defined in Provisions 
Sec. 4.1.2; SDS = (2/3) SMS, spectral acceleration in the short-period range for the design ground 
motions; and SD1 = (2/3) SM1, spectral acceleration at 1.0 second period for the design ground 
motions. 
As noted above, the design ground motions SDS and SD1 are defined as 2/3 times the maximum 
considered earthquake ground motions. The 2/3 factor is based on the estimated seismic margins in 
the design process of the Provisions as previously discussed (i.e., the design level of ground motion is 
1/1.5 or 2/3 times the maximum considered earthquake ground motion). 
Based on the above defined ground motions, the base shear is: 
V =CsW 
where DS 
S 
C RS I 
= and SDS = the design spectral response acceleration in the short period range as 
determined from Sec. 4.1.2, R = the response modification factor from Table 5.2.2, and I = the 
occupancy importance factor determined in accordance with Sec. 1.4. 
The value of Cs need not exceed ( ) 
Dl 
S 
C T SR I 
= but shall not be taken less than CS= 0.1SDl or, for 
buildings and structures in Seismic Design Categories E and F, 0.5 1 
S 
C R SI 
= 
where I and R are as defined above and SD1 = the design spectral response acceleration at a period of 
1.0 second as determined from Sec. 4.1.2, T = the fundamental period of the structure (sec) 
determined in Sec. 5.4.2, and S1 = the mapped maximum considered earthquake spectral response 
acceleration determined in accordance with Sec. 4.1. 
Where a design response spectrum is required by these Provisions and site-specific procedures are not 
used, the design response spectrum curve shall be developed as indicated in Figure A4 and as follows: 
1. For periods less than or equal to T0, the design spectral response acceleration, Sa, shall 
be taken as given by Eq. 4.1.2.6-1: 
0 
0.6 0.4 DS 
a DS 
S S T S 
T 
= + (4.1.2.6-1) 
2. For periods greater than or equal to T0 and less than or equal to TS, the design spectral response 
3. For periods greater than TS, the design spectral response acceleration, Sa, shall be taken as given 
by Eq. 4.1.2.6-3. 
Development of Maximum Considered Earthquake Ground Motion Maps Figures 3.3-1 through 3.314 
329 
Dl 
a 
S ST 
= (4.1.2.6-3) 
where: 
SDS = the design spectral response acceleration at short periods; 
SD1 = the design spectral response acceleration at 1 second period; 
T = the fundamental period of the structure (sec); 
T0 = 0.2SD1/SDS; and 
TS = SD1/SDS. 
Site-specific procedures for determining ground motions and response spectra are discussed in Sec. 
4.1.3 of the Provisions. 
REFERENCES 
Applied Technology Council. 1978. Tentative Provisions for the Development of Seismic Regulations 
for Building, Report ATC-3-06. Redwood City, California: Applied Technology Council. 
Structural Engineers Association of California. 1995. Vision 2000, Performance Based Seismic 
Engineering of Buildings. Prepared by the Structural Engineers Association of California, Vision 
2000 Committee, for the California Office of Emergency Services. 
Kennedy, R. P. and S. A. Short. 1994. Basis for Seismic Provisions of DOE-STD-1020, UCRL-CR- 
111478. Prepared for Lawrence Livermore National Laboratory. Livermore: University of California. 
FIGURE A4 Design response spectrum. 
2003 Commentary, Appendix A 
330 
Ellingwood, B. R. 1994. Probability-based Codified Design for Earthquakes. Engineering 
Structures 17(7). 
Cornell, C. A. 1994. Risk-Based Structural Design, Proceedings of Symposium on Risk Analysis, Ann 
Arbor: University of Michigan. 
Dewey, J. W., B. G. Reagor, L. Dengler, and K. Moley. 1995. Intensity Distribution and Isoseismal 
Maps for the Northridge, California, Earthquake of January 17, 1994, U. S. Geological Survey Open- 
File Report 95-92. 
Sommerville, P. 1995. Smooth Site Response Spectra Contours Developed for ATC-33," Appendix 
B of Guidelines for Seismic Rehabilitation of Existing Buildings, Northridge Earthquake Case-Study 
Report, ATC-33-1. Redwood City, California: Applied Technology Council. 
Frankel, et al. 1996. National Seismic-Hazard Maps: Documentation, June 1996, U.S. Geological 
Survey Open-File Report 96-532. 

 331 
Commentary Appendix B 
DEVELOPMENT OF THE USGS SEISMIC MAPS 
INTRODUCTION 
The 1997 Provisions used new design procedures based on the use of spectral response acceleration 
rather than the traditional peak ground acceleration and/or peak ground velocity, and these procedures 
are used again in the 2000 Provisions. The use of spectral ordinates and their relationship to building 
codes has been described by Leyendecker et al. (1995). The spectral response accelerations used in 
the new design approach are obtained from combining probabilistic maps (Frankel et al, 1996) 
prepared by the U.S. Geological Survey (USGS) with deterministic maps using procedures developed 
by the Building Seismic Safety Councils Seismic Design Procedures Group (SDPG). The SDPG 
recommendations are based on using the 1996 USGS probabilistic hazard maps with additional 
modifications based on review by the SDPG and the application of engineering judgment. This 
appendix summarizes the development of the USGS maps and describes how the 1997 and 2000 
Provisions design maps were prepared from them using SDPG recommendations. The SDPG effort 
has sometimes been referred to as Project >97. 
DEVELOPMENT OF PROBABILISTIC MAPS FOR THE UNITED STATES 
New seismic hazard maps for the conterminous United States were completed by the USGS in June 
1996 and placed on the World Wide Web (http://geohazards.cr.usgs.gov/eq/). The color maps can be 
viewed on the Web and/or downloaded to the users computer for printing. Paper copies of the maps 
are also available (Frankel et al, 1997a, 1997b). 
New seismic hazard maps for Alaska were completed by the USGS in January 1998 and placed on the 
USGS Web site (http://geohazards.cr.usgs.gov/eq/). Both documentation and printing of the maps are 
in progress (U. S. Geological Survey, 1998a, 1998b). 
New probabilistic maps are in preparation for Hawaii using the methodology similar to that used for 
the rest of the United States, and described below. These maps were to have been completed in early 
1998. Probabilistic maps for Puerto Rico, Culebra, Vieques, St. Thomas, St. John, St. Croix, Guam, 
and Tutuila needed for the 1997 Provisions are not expected during the current cycle of USGS map 
revisions (development of design maps for these areas is described below). 
This appendix provides a brief description of the USGS seismic hazard maps, the 
geologic/seismologic inputs to these maps, and the ground-motion relations used for the maps. It is 
based on the USGS map documentation for the central and eastern United States (CEUS) and the 
western United States (WUS) prepared by Frankel et al. (1996). The complete reference document, 
also available on the USGS Web site, should be reviewed for detailed technical information. 
The hazard maps depict probabilistic ground acceleration and spectral response acceleration with 10 
percent, 5 percent, and 2 percent probabilities of exceedance (PE) in 50 years. These maps correspond 
to return times of approximately 500, 1000, and 2500 years, respectively.1 All spectral response 
values shown in the maps correspond to 5 percent of critical damping. The maps are based on the 
assumption that earthquake occurrence is Poissonian, so that the probability of occurrence is time- 
1 Previous USGS maps (e.g. Algermissen et al., 1990 and Leyendecker et al., 1995) and earlier editions of the Provisions 
expressed probability as a 10 percent probability of exceedance in a specified exposure time. Beginning with the 1996 maps, 
probability is being expressed as a specified probability of exceedance in a 50 year time period. Thus, 5 percent in 50 years 
and 2 percent in 50 years used now correspond closely to 10 percent in 100 years and 10 percent in 250 years, respectively, 
that was used previously. This same information may be conveyed as annual frequency. In this approach 10 percent 
probability of exceedance (PE) in 50 years corresponds to an annual frequency of exceedance of 0.0021; 5 percent PE in 100 
years corresponds to 0.00103; and 2 percent PE in 50 years corresponds to 0.000404. 
2003 Commentary, Appendix B 
332 
independent. The methodologies used for the maps were presented, discussed, and substantially 
modified during 6 regional workshops for the conterminous United States convened by the USGS 
from June 1994-June 1995. A seventh workshop for Alaska was held in September 1996. 
The methodology for the maps (Frankel et al., 1996) includes three primary features: 
1. The use of smoothed historical seismicity is one component of the hazard calculation. This is used 
in lieu of source zones used in previous USGS maps. The analytical procedure is described in 
Frankel (1995). 
2. Another important feature is the use of alternative models of seismic hazard in a logic tree 
formalism. For the CEUS, different models based on different reference magnitudes are combined 
to form the hazard maps. In addition, large background zones based on broad geologic criteria are 
used as alternative source models for the CEUS and the WUS. These background zones are meant 
to quantify hazard in areas with little historic seismicity, but with the potential to produce major 
earthquakes. The background zones were developed from extensive discussions at the regional 
workshops. 
3. For the WUS, a big advance in the new maps is the use of geologic slip rates to determine fault 
recurrence times. Slip rates from about 500 faults or fault segments were used in preparing the 
probabilistic maps. 
The hazard maps do not consider the uncertainty in seismicity or fault parameters. Preferred values of 
maximum magnitudes and slip rates were used instead. The next stage of this effort is the 
quantification of uncertainties in hazard curves for selected sites. These data will be included on the 
Internet as they become available. 
The USGS hazard maps are not meant to be used for Mexico, areas north of 49 degrees north latitude, 
and offshore the Atlantic and Gulf of Mexico coasts of the United States. 
CEUS and WUS attenuation boundary. Attenuation of ground motion differs between the CEUS 
and the WUS. The boundary between regions was located along the eastern edge of the Basin and 
Range province (Figure B1). The previous USGS maps (e.g., Algermissen et al., 1990) used an 
attenuation boundary further to the east along the Rocky Mountain front. 
Separate hazard calculations were done for the two regions using different attenuation relations. 
Earthquakes west of the boundary used the WUS attenuation relations and earthquakes east of the 
boundary used CEUS attenuation relations. WUS attenuation relations were used for WUS 
earthquakes, even for sites located east of the attenuation boundary. Similarly CEUS attenuations 
were used for CEUS earthquakes, even for sites located west of the attenuation boundary. It would 
have been computationally difficult to consider how much of the path was contained in the attenuation 
province. Also, since the attenuation relation is dependent on the stress drop, basing the relation that 
was used on the location of the earthquake rather than the receiver is reasonable. 
Hazard curves. The probabilistic maps were constructed from mean hazard curves, that is the mean 
probabilities of exceedance as a function of ground motion or spectral response. Hazard curves were 
obtained for each site on a calculation grid. 
A grid (or site) spacing of 0.1 degrees in latitude and longitude was used for the WUS and 0.2 degrees 
for the CEUS. This resulted in hazard calculations at about 65,000 sites for the WUS runs and 35,000 
sites for the CEUS runs. The CEUS hazard curves were interpolated to yield a set of hazard curves on 
a 0.1 degree grid. A grid of hazard curves with 0.1 degree spacing was thereby obtained for the entire 
conterminous United States. A special grid spacing of 0.05 degrees was also done for California, 
Nevada, and western Utah because of the density of faults warranted increased density of data. These 
data were used for maps of this region. 

Development of the USGS Seismic Maps 
333 
Figure B1 Attenuation boundary for eastern and western attenuation function. 
Figure B2 is a sample of mean hazard curves used in making the 1996 maps. The curves include cities 
from various regions in the United States. It should be noted that in some areas the curves are very 
sensitive to the latitude and longitude selected. A probabilistic map is a contour plot of the ground 
motion or spectral values obtained by taking a slice through all 150,000 hazard curves at a particular 
probability value. The gridded data obtained from the hazard curves that was used to make each 
probabilistic map is located at the USGS Web site. Figure B2 also shows the general difference in 
slope of the hazard curves of the CEUS versus the WUS. This difference has been noted in other 
studies. 

2003 Commentary, Appendix B 
334 
CENTRAL AND EASTERN UNITED STATES 
The basic procedure for constructing the CEUS portion of the hazard maps is diagrammed in Figure 
B3. Four models of hazard are shown on the left side of the figure. Model 1 is based on mb 3.0 and 
larger earthquakes since 1924. Model 
2 is derived from mb 4.0 and larger 
earthquakes since 1860. Model 3 is 
produced from mb 5.0 and larger 
events since 1700. In constructing 
the hazard maps, model 1 was 
assigned a weight twice that of 
models 2 and 3. 
The procedure described by Frankel 
(1995) is used to construct the hazard 
maps directly from the historic 
seismicity (models 1-3). The number 
of events greater than the minimum 
magnitude are counted on a grid with 
spacing of 0.1 degrees in latitude and 
longitude. The logarithm of this 
number represents the maximum 
likelihood a-value for each grid cell. 
Note that the maximum likelihood 
method counts a mb 5 event the same 
as a mb 3 event in the determination 
of a-value. Then the gridded a-values 
are smoothed using a Gaussian 
function. A Gaussian with a 
correlation distance of 50 km was 
used for model 1 and 75 km for 
models 2 and 3. The 50 km distance 
was chosen because it is similar in 
width to many of the trends in historic seismicity in the CEUS. In addition, it is comparable to the 
Figure B3 Seismic hazard models for the central and 
eastern United States. Smoothed seismicity models are 
shown on the left and fault models are shown on the right. 
Development of the USGS Seismic Maps 
335 
error in location of mb 3 events in the period of 1924-1975, before the advent of local seismic 
networks. A larger correlation distance was used for models 2 and 3 since they include earthquakes 
further back in time with poorer estimates of locations. 
Model 4 consists of large background source zones. This alternative is meant to quantify hazard in 
areas with little historical seismicity but with the potential to generate damaging earthquakes. These 
background zones are detailed in a later section of this text. The sum of the weights of models 1-4 is 
one. For a weighting scheme that is uniform in space, this ensures that the total seismicity rate in the 
combined model equals the historic seismicity rate. A spatially-varying weighting scheme which 
slightly exceeds the historic seismicity rate was used in the final map for reasons which are described 
later. 
A regional b-value of 0.95 was used for models 1-4 in all of the CEUS except Charlevoix, Quebec. 
This b-value was determined from a catalog for events east of 105 degrees W. For the Charlevoix 
region a b-value of 0.76 was used based on the work of John Adams, Stephen Halchuck and Dieter 
Weichert of the Geologic Survey of Canada (see Adams et al., 1996). 
Figure B4 shows a map of the CEUS Mmax values used for models 1-4 (bold M refers to moment 
magnitude). These Mmax zones correspond to the background zones used in model 4. Most of the 
CEUS is divided into a cratonic region and a region of extended crust. An Mmax of 6.5 was used for 
the cratonic area. A Mmax of 7.5 was used for the Wabash Valley zone in keeping with magnitudes 
derived from paleoliquefaction evidence (Obermeier et al., 1992). An Mmax of 7.5 was used in the 
zone of extended crust outboard of the craton. An Mmax of 6.5 was used for the Rocky Mountain zone 
and the Colorado Plateau, consistent with the magnitude of the largest historic events in these regions. 
An Mmax of 7.2 was used for the gridded seismicity within the Charleston areal source zone. A 
minimum mb of 5.0 was used in all the hazard calculations for the CEUS. 
Figure B4 Central and eastern U.S. maximum magnitude zones. 
Model 5 (Figure B3, right) consists of the contribution from large earthquakes (M > 7.0) in four 
specific areas of the CEUS: New Madrid, Charleston, South Carolina, the Meers fault in southwest 
Oklahoma, and the Cheraw Fault in eastern Colorado. This model has a weight of 1. The treatment of 
2003 Commentary, Appendix B 
336 
these special areas is described below. There are three other areas in the CEUS that are called special 
zones: eastern Tennessee, Wabash Valley, and Charlevoix. These are described below. 
Special zones A number of special case need to be described. 
New Madrid: To calculate the hazard from large events in the New Madrid area, three parallel faults 
in an S-shaped pattern encompassing the area of highest historic seismicity were considered. These 
were not meant to be actual faults; they are simply a way of expressing the uncertainty in the source 
locations of large earthquakes such as the 1811-12 sequence. A characteristic rupture model with a 
characteristic moment magnitude M of 8.0, similar to the estimated magnitudes of the largest events in 
1811-12 (Johnston, 1996a, 1996b) was assumed. A recurrence time of 1000 years for such an event 
was used as an average value, considering the uncertainty in the magnitudes of pre-historic events. 
An areal source zone was used for New Madrid for models 1-3, rather than spatially-smoothed historic 
seismicity. This zone accounts for the hazard from New Madrid events with moment magnitudes less 
than 7.5. 
Charleston, South Carolina: An areal source zone was used to quantify the hazard from large 
earthquakes. The extent of the areal source zone was constrained by the areal distribution of 
paleoliquefaction locations, although the source zone does not encompass all the paleoliquefaction 
sites. A characteristic rupture model of moment magnitude 7.3 earthquakes, based on the estimated 
magnitude of the 1886 event (Johnston, 1996b) was assumed. For the M7.3 events a recurrence time 
of 650 years was used, based on dates of paleoliquefaction events (Amick and Gelinas, 1991; 
Obermeier et al., 1990; Johnston and Schweig, written comm., 1996). 
Meers Fault: The Meers fault in southwestern Oklahoma was explicitly included. The segment of the 
fault which has produced a Holocene scarp as described in Crone and Luza (1990) was used. A 
characteristic moment magnitude of 7.0 and a recurrence time of 4000 years was used based on their 
work. 
Cheraw Fault: This eastern Colorado fault with Holocene faulting based on a study by Crone et al. 
(1996) was included. The recurrence rate of this fault was obtained from a slip rate of 0.5 mm/yr. A 
maximum magnitude of 7.1 was found from the fault length using the relations of Wells and 
Coppersmith (1994). 
Eastern Tennessee Seismic Zone: The eastern Tennessee seismic zone is a linear trend of seismicity 
that is most obvious for smaller events with magnitudes around 2 (see Powell et al., 1994). The 
magnitude 3 and larger earthquakes tend to cluster in one part of this linear trend, so that hazard maps 
are based just on smoothed mb3. 
Wabash Valley: Recent work has identified several paleoearthquakes in the areas of southern Indiana 
and Illinois based on widespread paleoliquefaction features (Obermeier et al., 1992). An areal zone 
was used with a higher Mmax of 7.5 to account for such large events. The sum of the gridded a-values 
in this zone calculated from model 1 produce a recurrence time of 2600 years for events with mb 6.5. 
The recurrence rate of M6.5 and greater events is estimated to be about 4,000 years from the 
paleoliquefaction dates (P. Munson and S. Obermeier, pers. comm., 1995), so it is not necessary to add 
additional large events to augment models 1-3. The Wabash Valley Mmax zone in the maps is based on 
the Wabash Valley fault zone. 
Charlevoix, Quebec: As mentioned above, a 40 km by 70 km region surrounding this seismicity 
cluster was assigned a b-value of 0.76, based on the work of Adams, Halchuck and Weichert. This bvalue 
was used in models 1-3. 
Background source zones (Model 4). The background source zones (see Figure B5) are intended to 
quantify seismic hazard in areas that have not had significant historic seismicity, but could very well 
produce sizeable earthquakes in the future. They consist of a cratonic zone, an extended margin zone, 
a Rocky Mountain zone, and a Colorado Plateau zone. The Rocky Mountain zone was not discussed 
at any workshop, but is clearly defined by the Rocky Mountain front on the east and the areas of 
Development of the USGS Seismic Maps 
337 
extensional tectonics to the west, north and south. As stated above, the dividing line between the 
cratonic and extended margin zone was drawn by Rus Wheeler based on the westward and northern 
edge of rifting during the opening of the Iapetan ocean. One justification for having craton and 
extended crust zones is the work done by Johnston (1994). They compiled a global survey of 
earthquakes in cratonic and extended crust and found a higher seismicity rate (normalized by area) for 
the extended areas. 
For each background zone, a-values were determined by counting the number of mb3 and larger events 
within the zone since 1924 and adjusting the rate to equal that since 1976. A b-value of 0.95 was used 
for all the background zones, based on the b-value found for the entire CEUS. 
Adaptive weighting for CEUS. The inclusion of background zones lowers the probabilistic ground 
motions in areas of relatively high historic seismicity while raising the hazard to only low levels in 
areas with no historic seismicity. The June 1996 versions of the maps include the background zones 
using a weighting scheme that can vary locally depending on the level of historic seismicity in that cell 
of the a-value grid. Spatially-varying weighting was suggested by Allin Cornell in the external review 
of the interim maps. The adaptive weighting procedure avoids lowering the hazard in higher 
seismicity areas to raise the hazard in low seismicity areas. This was implemented by looping through 
the a-value grid and checking to see if the a-value for each cell from the historic seismicity was greater 
than the a-value from the background zone. For the CEUS the a-value from the historic seismicity 
was derived by weighting the rates from models 1, 2, and 3 by 0.5, 0.25, 0.25 respectively. If this 
weighted sum was greater than the rate from the appropriate background zone, then the rate for that 
cell was determined by weighting the rates from models 1-3 by 0.5, .25, .25 (i.e., historic seismicity 
only, no background zone). If the weighted sum from the historic seismicity was less than the rate of 
the background zone, then a weighting of 0.4, 0.2, 0.2, 0.2 for models 1-4, respectively (including the 
background zone as model 4). This procedure does not make the rate for any cell lower than it would 
be from the historic seismicity (models 1-3). It also incorporates the background zones in areas of low 
historic seismicity. The total seismicity rate in the resulting a-value grid is only 10 percent larger than 
the observed rate of mb3=s since 1976. This is not a major difference. Of course, this procedure 
produces substantially higher ground motions (in terms of percentage increase) in the seismically quiet 
areas as compared to no background zone. These values are still quite low in an absolute sense. 

2003 Commentary, Appendix B 
338 
Figure B5 Central and eastern U.S. background zones. 
CEUS catalogs and b-value calculation. The primary catalog used for the CEUS for longitudes east 
of 105 degrees is Seeber and Armbruster (1991), which is a refinement of the EPRI (1986) catalog. 
This was supplemented with the PDE catalog from 1985-1995. In addition, PDE, DNAG, Stover and 
Coffman (1993), Stover, Reagor, and Algermissen (1984) catalogs were searched to find events not 
included in Seeber and Armbruster (1991). Mueller et al. (1996) describes the treatment of catalogs, 
adjustment of rates to correct for incompleteness, the removal of aftershocks, and the assignment of 
magnitudes. 
Attenuation relations for CEUS. The reference site condition used for the maps is specified to be 
the boundary between Provisions Site Classes B and C (Martin and Dobry, 1994), meaning it has an 
average shear-wave velocity of 760 m/sec in the top 30m. This corresponds to a typical firm-rock 
site for the western United States (see WUS attenuation section below), although many rock sites in 
the CEUS probably have much higher velocities. The motivation for using this reference site is that it 
corresponds to the average of sites classified as rock sites in WUS attenuation relations. In addition, 
it was considered less problematic to use this site condition for the CEUS than to use a soil condition. 
Most previously-published attenuation relations for the CEUS are based on a hard-rock site condition. 
It is less of a problem to convert these to a firm-rock condition than to convert them to a soil 
condition, since there would be less concern over possible non-linearity for the firm-rock site 
compared to the soil site. 
Two equally-weighted, attenuation relations were used for the CEUS. Both sets of relations were 
derived by stochastic simulations and random vibration theory. First the Toro et al. (1993) attenuation 
for hard-rock was used. The attenuation relations were multiplied by frequency-dependent factors 
developed by USGS to convert them from hard-rock to firm-rock sites. The factors used 1.52 for 
PGA, 1.76 for 0.2 sec spectral response, 1.72 for 0.3 sec spectral response and 1.34 for 1.0 sec spectral 
response. These factors were applied independently of magnitude and distance. 
The second set of relations was derived by USGS (Frankel et al., 1996) for firm-rock sites. These 
relations were based on a Brune source model with a stress drop of 150 bars. The simulations 
contained frequency-dependent amplification factors derived from a hypothesized shear-wave velocity 
Development of the USGS Seismic Maps 
339 
profile of a CEUS firm-rock site. A series of tables of ground motions and response spectral values as 
a function of moment magnitude and distance was produced instead of an equation. 
For CEUS hazard calculations for models 1-4, a source depth of 5.0 km was assumed when using the 
USGS ground motion tables. Since a minimum hypocentral distance of 10 km is used in the USGS 
tables, the probabilistic ground motions are insensitive to the choice of source depth. In the hazard 
program, when hypocentral distances are less than 10 km the distance is set to 10 km when using the 
tables. For the Toro et al. (1993) relations, the fictitious depths that they specify for each period are 
used, so that the choice of source depth used in the USGS tables was not applied. 
For both sets of ground motion relations, values of 0.75, 0.75, 0.75, and 0.80 were used for the natural 
logarithms of the standard deviation of PGA, 0.2 sec, 0.3 sec, and 1.0 sec spectral responses, 
respectively. These values are similar to the aleatory standard deviations reported to the Senior 
Seismic Hazard Analysis Committee (1996). 
A cap in the median ground motions was placed on the ground motions within the hazard code. USGS 
was concerned that the median ground motions of both the Toro et al. and the new USGS tables 
became very large (> 2.5g PGA) for distances of about 10 km for the M 8.0 events for New Madrid. 
Accordingly the median PGAs was capped at 1.5g. The median 0.3 and 0.2 sec values were capped 
at 3.75g which was derived by multiplying the PGA cap by 2.5 (the WUS conversion factor). This 
only affected the PGA values for the 2 percent PE in 50 year maps for the area directly above the three 
fictitious faults for the New Madrid region. It does not change any of the values at Memphis. The 
capping did not significantly alter the 0.3 and 0.2 sec values in this area. The PGA and spectral 
response values did not change in the Charleston region from this capping. Note that the capping was 
for the median values only. As the variability (sigma) of the ground motions was maintained in the 
hazard code, values larger than the median were allowed. USGS felt that the capping recognizes that 
values derived from point source simulations are not as reliable for M8.0 earthquakes at close-in 
distances (< 20 km). 
Additional notes for CEUS. One of the major outcomes of the new maps for the CEUS is that the 
ground motions are about a factor of 2 to 3 times lower, on average, than the PGA values in 
Algermissen et al. (1990) and the spectral values in Algermissen et al. (1991) and Leyendecker et al. 
(1995). The primary cause of this difference is the magnitudes assigned to pre-instrumental 
earthquakes in the catalog. Magnitudes of historic events used by Algermissen et al. were based on 
Imax (maximum observed intensity), using magnitude-Imax relations derived from WUS earthquakes. 
This overestimates the magnitudes of these events and, in turn, overestimates the rates of M4.9 and 
larger events. The magnitudes of historic events used in the new maps were primarily derived by 
Seeber and Armbruster (1991) from either felt area or Imax using relations derived from CEUS 
earthquakes (Sibol et al., 1987). Thus, rates of M4.9 and larger events are much lower in the new 
catalog, compared to those used for the previous USGS maps. 
It is useful to compare the new maps to the source zones used in the EPRI (1986) study. For the areas 
to the north and west of New Madrid, most of the six EPRI teams had three source zones in common: 
1) the Nemaha Ridge in Kansas and Nebraska, 2) the Colorado-Great Lakes lineament extending from 
Colorado to the western end of Lake Superior, and 3) a small fault zone in northern Illinois, west of 
Chicago. Each of these source zones are apparent as higher hazard areas in the our maps. The 
Nemaha Ridge is outlined in the maps because of magnitude 4 and 5 events occurring in the vicinity. 
Portions of the Colorado-Great Lakes lineament show higher hazard in the map, particularly the 
portion in South Dakota and western Minnesota. The portion of the lineament in eastern Minnesota 
has been historically inactive, so is not apparent on the maps. The area in western Minnesota shows 
some hazard because of the occurrence of a few magnitude 4 events since 1860. A recent paper by 
Chandler (1995), argues that the locations and focal mechanisms of these earthquakes are not 
compatible with them being on the lineament, which is expressed as the Morris Fault in this region. 
The area in northern Illinois has relatively high hazard in the maps because of M4-5 events that have 
occurred there. 
2003 Commentary, Appendix B 
340 
Frankel (1995) also found good agreement in the mean PEs and hazard curves derived from models 
1-3 and 4 and those produced by the EPRI (1986) study, when the same PGA attenuation relations 
were used. 
WESTERN UNITED STATES 
The maps for the WUS include a cooperative effort with the California Division of Mines and 
Geology. This was made possible, in part, because CDMG was doing a probabilistic map at the same 
time the USGS maps were prepared. There was considerable cooperation in this effort. For example, 
the fault data base used in the USGS maps was obtained from CDMG. Similarly USGS software was 
made available to CDMG. The result 
is that maps produced by both 
agencies are the same. 
The procedure for mapping hazard in 
the WUS is shown in Figure B6. On 
the left side, hazards are considered 
from earthquakes with magnitudes less 
than or equal to moment magnitude 
7.0. For most of the WUS, two 
alternative models are used: 1) 
smoothed historical seismicity (weight 
of 0.67) and 2) large background 
zones (weight 0.33) based on broad 
geologic criteria and workshop input. 
Model 1 used a 0.1 degree source grid 
to count number of events. The 
determination of a-value was changed 
somewhat from the CEUS, to 
incorporate different completeness 
times for different magnitude ranges. 
The a-value for each grid cell was 
calculated from the maximum 
likelihood method of Weichert (1980), 
based on events with magnitudes of 
4.0 and larger. The ranges used were M4.0 to 5.0 since 1963, M5.0 to 6.0 since 1930, and M6.0 and 
larger since 1850. For the first two categories, completeness time was derived from plots of 
cumulative number of events versus time. M3 events were not used in the WUS hazard calculations 
since they are only complete since about 1976 for most areas and may not even be complete after 1976 
for some areas. For California M4.0 to M5.0 since 1933, M5.0 to 6.0 since 1900, and M6.0 and larger 
since 1850 were used. The catalog for California is complete to earlier dates compared to the catalogs 
for the rest of the WUS (see below). 
Another difference with the CEUS is that multiple models with different minimum magnitudes for the 
a-value estimates (such as models 1-3 for the CEUS) were not used. The use of such multiple models 
in the CEUS was partially motivated by the observation that some mb4 and mb5 events in the CEUS 
occurred in areas with few mb3 events since 1924 (e.g., Nemaha Ridge events and western Minnesota 
events). It was considered desirable to be able to give such mb4 and mb5 events extra weight in the 
hazard calculation over what they would have in one run with a minimum magnitude of 3. In contrast 
it appears that virtually all M5 and M6 events in the WUS have occurred in areas with numerous M4 
events since 1965. There was also reluctance to use a WUS model with a-values based on a minimum 
magnitude of 6.0, since this would tend to double count events that have occurred on mapped faults 
included in Figure B6 right. 
Figure B6 Seismic hazard models for California and the 
western United States. Smoothed seismicity models are 
shown on the left and fault models are shown on the right.
Development of the USGS Seismic Maps 
341 
For model 1, the gridded a-values were smoothed with a Gaussian with a correlation distance of 50 
km, as in model 1 for the CEUS. The hazard calculation from the gridded a-values differed from that 
in the CEUS, because we considered fault finiteness in the WUS calculations. For each source grid 
cell, a fictitious fault for magnitudes of 6.0 and larger was used. The fault was centered on the center 
of the grid cell. The strike of the fault was random and was varied for each magnitude increment. The 
length of the fault was determined from the relations of Wells and Coppersmith (1994). The fictitious 
faults were taken to be vertical. 
A maximum moment magnitude of 7.0 was used for models 1 and 2, except for four shear zones in 
northeastern California and western Nevada described below. Of course, larger moment magnitudes 
are included in the specific faults. A minimum moment magnitude of 5.0 were used for models 1 and 
2. For each WUS site, the hazard calculation was done for source-site distances of 200 km and less, 
except for the Cascadia subduction zone, where the maximum distance was 1000 km. 
Separate hazard calculations for deep events (> 35 km) were done. These events were culled from the 
catalogs. Their a-values were calculated separately from the shallow events. Different attenuation 
relations were used. 
Regional b-values were calculated based on the method of Weichert (1980), using events with 
magnitudes of 4 and larger and using varying completeness times for different magnitudes. 
Accordingly, a regional b-value of 0.8 was used in models 1 and 2 for the WUS runs based on shallow 
events. For the deep events (> 35 km), an average b-value of 0.65 was found. This low b-value was 
used in the hazard calculations for the deep events. 
We used a b-value of 0.9 for most of California, except for the easternmost portion of California in our 
basin and range background zone (see below). This b-value was derived by CDMG. 
Faults. The hazard from about 500 Quaternary faults or fault segments was used for the maps. Faults 
were considered where geologic slip rates have been determined or estimates of recurrence times have 
been made from trenching studies. A table of the fault parameters used in the hazard calculations has 
been compiled and is shown on the USGS Internet Web site. Figure B7 shows the faults used in the 
maps. The numerous individuals who worked on compilations of fault data are too numerous to cite 
here. They are cited, along with their contribution, in the map documentation (Frankel et al, 1996). 
Recurrence models for faults. The hazard from specific faults is added to the hazard from the 
seismicity as shown in Figure B6. Faults are divided into types A and B, roughly following the 
nomenclature of WGCEP (1995). A fault is classified as A-type if there have been sufficient studies 
of it to produce models of fault segmentation. In California the A-type faults are: San Andreas, San 
Jacinto, Elsinore, Hayward, Rodgers Creek, and Imperial (M. Petersen, C. Cramer, and W. Bryant, 
written comm., 1996). The only fault outside of California classified as an A-type is the Wasatch 
Fault. Single-segment ruptures were assumed on the Wasatch Fault. 
2003 Commentary, Appendix B 
342 
Figure B7 Western U.S. faults included in the maps. 
For California, the rupture scenarios specified by Petersen, Cramer and Bryant of CDMG, with input 
from Lienkaemper of USGS for northern California were used. Single-segment, characteristic rupture 
for the San Jacinto and Elsinore faults were assumed. For the San Andreas fault, multiple-segment 
ruptures were included in the hazard calculation, including repeats of the 1906 and 1857 rupture zones, 
and a scenario with the southern San Andreas fault rupturing from San Bernardino through the 
Coachella segment. Both single-segment and double-segment ruptures of the Hayward Fault were 
included. 
For California faults, characteristic magnitudes derived by CDMG from the fault area using the 
relations in Wells and Coppersmith (1994) were used. For the remainder of the WUS, the 
characteristic magnitude was determined from the fault length using the relations of Wells and 
Coppersmith (1994) appropriate for that fault type. 
Development of the USGS Seismic Maps 
343 
For the B-type faults, it was felt there were insufficient studies to warrant specific segmentation 
boundaries. For these faults, the scheme of Petersen et al. (1996) was followed, using both 
characteristic and Gutenberg-Richter (G-R; exponential) models of earthquake occurrence. These 
recurrence models were weighted equally. The G-R model basically accounts for the possibility that a 
fault is segmented and may rupture only part of its length. It was assumed that the G-R distribution 
applies from a minimum moment magnitude of 6.5 up to a moment magnitude corresponding to 
rupture of the entire fault length. 
The procedure for calculating hazard using the G-R model involves looping through magnitude 
increments. For each magnitude a rupture length is calculated using Wells and Coppersmith (1994). 
Then a rupture zone of this length is floated along the fault trace. For each site, the appropriate 
distance to the floating ruptures is found and the frequency of exceedance (FE) is calculated. The 
FEs are then added for all the floating rupture zones. 
As used by USGS, the characteristic earthquake model (Schwartz and Coppersmith, 1984) is actually 
the maximum magnitude model of Wesnousky (1986). Here it is assumed that the fault only generates 
earthquakes that rupture the entire fault. Smaller events along the fault would be incorporated by 
models 1 and 2 with the distributed seismicity or by the G-R model described above. 
It should be noted that using the G-R model generally produces higher probabilistic ground motions 
than the characteristic earthquake model, because of the more frequent occurrence of earthquakes with 
magnitudes of about 6.5. 
Fault widths (except for California) were determined by assuming a seismogenic depth of 15 km and 
then using the dip, so that the width equaled 15 km divided by the sine of the dip. For most normal 
faults a dip of 60 degrees is assumed. Dip directions were taken from the literature. For the Wasatch, 
Lost River, Beaverhead, Lemhi, and Hebgen Lake faults, the dip angles were taken from the literature 
(see fault parameter table on Web site). Strike-slip faults were assigned a dip of 90 degrees. For 
California faults, widths were often defined using the depth of seismicity (J. Lienkaemper, written 
comm., 1996; M. Petersen, C. Cramer, and W. Bryant, written comm., 1996). Fault length was 
calculated from the total length of the digitized fault trace. 
Special cases. There are a number of special cases which need to be described. 
Blind thrusts in the Los Angeles area: Following Petersen et al. (1996) and as discussed at the 
Pasadena workshop, 0.5 weight was assigned to blind thrusts in the L.A. region, because of the 
uncertainty in their slip rates and in whether they were indeed seismically active. These faults are the 
Elysian Park thrust and the Compton thrust. The Santa Barbara Channel thrust (Shaw and Suppe, 
1994) also has partial weight, based on the weighting scheme developed by CDMG. 
Offshore faults in Oregon: A weight of 0.05 was assigned to three offshore faults in Oregon identified 
by Goldfinger et al. (in press) and tabulated by Geomatrix (1995): the Wecoma, Daisy Bank and Alvin 
Canyon faults. It was felt the uncertainty in the seismic activity of these faults warranted a low 
weight, and the 0.05 probability of activity decided in Geomatrix (1995) was used. A 0.5 weight was 
assigned to the Cape Blanco blind thrust. 
Lost River, Lemhi and Beaverhead faults in Idaho: It was assumed that the magnitude of the Borah 
Peak event (M7.0) represented a maximum magnitude for these faults. As with (3), the characteristic 
model floated a M7.0 along each fault. The G-R model considered magnitudes between 6.5 and 7.0. 
Note that using a larger maximum magnitude would lower the probabilistic ground motions, because it 
would increase the recurrence time. 
Hurricane and Sevier-Torroweap Faults in Utah and Arizona: The long lengths of these faults (about 
250 km) implied a maximum magnitude too large compared to historical events in the region. 
Therefore a maximum magnitude of M7.5 was chosen. The characteristic and G-R models were 
implemented as in case (3). Other faults (outside of California) where the Mmax was determined to be 
greater than 7.5 based on the fault length were assigned a maximum magnitude of 7.5. 
2003 Commentary, Appendix B 
344 
Wasatch Fault in Utah: Recurrence times derived from dates of paleoearthquakes by Black et al. 
(1995) and the compilation of McCalpin and Nishenko (1996) were used 
Hebgen Lake Fault in Montana: A characteristic moment magnitude of 7.3 based on the 1959 event 
(Doser, 1985) was used. 
Short faults: All short faults with characteristic magnitudes of less than 6.5 were treated with the 
characteristic recurrence model only (weight = 1). No G-R relation was used. If a fault had a 
characteristic magnitude less than 6.0, it was not used. 
Seattle Fault: The characteristic recurrence time was fixed at 5000 years, which is the minimum 
recurrence time apparent from paleoseismology (R. Bucknam, pers. comm., 1996). Using the 
characteristic magnitude of 7.1 derived from the length and a 0.5 mm/yr slip rate yielded a 
characteristic recurrence time of about 3000 years. 
Eglington fault near Las Vegas: The recurrence time for this fault was fixed at 14,000 years, similar 
to the recurrence noted in Wyman et al. (1993). 
Shear Zones in Eastern California and Western Nevada: Areal shear zones were added along the 
western border of Nevada extending from the northern end of the Death Valley fault through the 
Tahoe-Reno area through northeast California ending at the latitude of Klamath Falls, Oregon. A 
shear rate of 4 mm/yr to zone 1, and 2 mm/yr to zones 2 and 3 was assigned. The shear rate in zone 1 
is comparable to the shear rate observed on the Death Valley fault, but which is not observed in 
mapped faults north of the Death Valley fault (C. dePolo and J. Anderson, pers. comm., 1996). For 
the Foothills Fault system (zone 4) a shear rate of 0.05 mm/yr was used. a-values were determined for 
these zones in the manner described in Ward(1994). For zones 1 through 3, a magnitude range of 6.5 
to 7.3 was used. For zone 4, a magnitude range of 6.0 to 7.0 was used. The maximum magnitude for 
the calculation of hazard from the smoothed historic seismicity was lowered in these zones so that it 
did not overlap with these magnitude ranges. Fictitious faults with a fixed strike were used in the 
hazard calculation for these zones. Again, use of these areal zones in California was agreed upon after 
consultation with CDMG personnel. 
Cascadia subduction zone. Two alternative scenarios for great earthquakes on the Cascadia 
subduction zone were considered. For both scenarios it was assumed that the recurrence time of 
rupture at any point along the subduction zone was 500 years. This time is in or near most of the 
average intervals estimated from coastal and offshore evidence (see Atwater and Hemphill-Haley, 
1996; Geomatrix, 1995; B. Atwater, written comm., 1996). Individual intervals, however, range from 
a few hundred years to about 1000 years (Atwater et al., 1995). 
The first scenario is for moment magnitude 8.3 earthquakes to fill the subduction zone every 500 
years. Based on a rupture length of 250 km (see Geomatrix, 1995) for an M8.3 event and the 1100 km 
length of the entire subduction zone, this requires a repeat time of about 110 years for an M8.3 event. 
However, no such event has been observed in the historic record of about 150 years. This M8.3 
scenario is similar to what was used in the 1994 edition of the USGS maps (see Leyendecker et al., 
1995) and it is comparable to the highest weighted scenario in Geomatrix (1995). A M8.3 rupture 
zone was floated along the strike of the subduction zone to calculate the hazard. A weight of 0.67 was 
assigned for this scenario in the maps. 
The second scenario used is for a moment magnitude 9.0 earthquake to rupture the entire Cascadia 
subduction zone every 500 years on average. No compelling reason was seen to rule out such a 
scenario. This scenario would explain the lack of M8 earthquakes in the historic record. It is also 
consistent with a recent interpretation of Japanese tsunami records by Satake et al. (1996). By ruling 
out alternative source regions, Satake et al. (1996) reported that a tsunami in 1700 could have been 
produced by a M9.0 earthquake along the Cascadia subduction zone. A weight of 0.33 was assigned 
to the M9.0 scenario in the maps. 
The subduction zone was specified as a dipping plane striking north-south from about Cape 
Mendocino to 50 degrees north. It was assumed that the plane reached 20 km depth at a longitude of 
Development of the USGS Seismic Maps 
345 
123.8 degrees west, just east of the coastline. This corresponds roughly to the 20 km depth contour 
drawn by Hyndman and Wang (1995) and is consistent with the depth and location of the Petrolia 
earthquake in northern California. A dip of 10 degrees was assigned to the plane and a width of 90 
km. The seismogenic portion of the plane was assumed to extend to a depth of 20 km. 
Background source zones. The background source zones for the WUS (model 2) were based on 
broad geologic criteria and were developed by discussion at the Salt Lake City (SLC) workshop 
(except for the Cascades source zone). These zones are shown in Figure B8. Note that there are no 
background source zones west of the Cascades and west of the Basin and Range province. For those 
areas, model 1 was used with a weight of 1. 
Figure B8 Western U.S. background zones. 
At the SLC workshop there was substantial sentiment for a Yellowstone Parabola source zone (see, 
e.g., Anders et al., 1989) that would join up seismically-active areas in western Wyoming with the 
source areas of the Bora Peak and Hebgen Lake earthquakes. It was felt that the relatively 
seismically-quiescent areas consisting of the Snake River Plain and Colorado Plateau should be 
separate source zones because of the geologic characteristics. An area of southwest Arizona was 
suggested as a separate source zone by Bruce Scheol, based partly on differences in the age and length 
of geologic structures compared with the Basin and Range Province (see Edge et al., 1992). A 
Cascades source zone was added since it was felt that was a geologically-distinct area. 
The remaining background source zone includes the Basin and Range Province, the Rio Grande Rift, 
areas of Arizona and New Mexico, portions of west Texas, and areas of eastern Washington and 
northern Idaho and Montana. The northern border of this zone follows the international border. As 
stated above, this seems to be a valid approach since the hazard maps are being based on the seismicity 
rate in the area of interest. 
This large background zone is intended to address the possibility of having large earthquakes (M6 and 
larger) in areas with relatively low rates of seismicity in the brief historic record. It is important to 
have a large zone that contains areas of high seismicity in order to quantify the hazard in relatively 
quiescent areas such as eastern Oregon and Washington, central Arizona, parts of New Mexico, and 
2003 Commentary, Appendix B 
346 
west Texas. One can see the effect of this large background zone by noting the contours on the hazard 
maps in these areas. The prominence of the background zones in the maps is determined by the 
weighting of models 1 and 2. 
Adaptive weighting for the WUS. The adaptive weighting procedure was used to include the 
background zones in the WUS without lowering the hazard values in the high seismicity areas. As 
with the CEUS, the a-value was checked for each source cell to see whether the rate from the historic 
seismicity exceeded that from the appropriate background zone. If it did, the a-value was used from 
the historic seismicity. If the historic seismicity a-value was below the background value, then a rate 
derived from using 0.67 times the historic rate plus 0.33 times the background rate was used. This 
does not lower the a-value in any cell lower than the value from the historic seismicity. The total 
seismicity rate in this portion of the WUS in the new a-value grid is 16 percent above the historic rate 
(derived from M4 and greater events since 1963). 
WUS catalogs. For the WUS, except for California, the Stover and Coffman (1993), Stover, Reagor, 
and Algermissen (1984), PDE, and DNAG catalogs (with the addition of Alan Sanfords catalog for 
New Mexico) were used. For California, a catalog compiled by Mark Petersen of California Division 
of Mines and Geology (CDMG) was used. Mueller et al. (1996) describes the processing of the 
catalogs, the removal of aftershocks, and the assignment of magnitudes. Utah coal-mining events 
were removed from the catalog (see Mueller et al., 1996). Explosions at NTS and their aftershocks 
were also removed from the catalog. 
Attenuation relations for WUS. These relations are discussed below. 
Crustal Events: For spectral response acceleration, three equally-weighted attenuation relations were 
used: (1) Boore, Joyner, and Fumal (BJF; 1993, 1994a) with later modifications to differentiate thrust 
and strike-slip faulting (Boore et al., 1994b) and (2) Sadigh et al. (1993). For (1) ground motions were 
calculated for a site with average shear-wave velocity of 760 m/sec in the top 30m, using the relations 
between shear-wave velocity and site amplification in Boore et al. (1994a). For (2) their rock values 
were used. Joyner (1995) reported velocity profiles compiled by W. Silva and by D. Boore showing 
that WUS rock sites basically spanned the NEHRP B/C boundary. When calculating ground motions 
for each fault, the relations appropriate for that fault type (e.g, thrust) were used. All of the relations 
found higher ground motions for thrust faults compared with strike slip faults. 
All calculations included the variability of ground motions. For 1) the sigma values reported in BJF 
(1994b) were used. For 2) the magnitude-dependent sigmas found in those studies were used. 
The distance measure from fault to site varies with the attenuation relation and this was accounted for 
in the hazard codes (see B.5 for additional detail on distance measures). 
Deep events (> 35 km): Most of these events occurred beneath the Puget Sound region, although 
some were in northwestern California. For these deep events, only one attenuation relation was used B 
that is, that developed by Geomatrix (1993; with recent modification for depth dependence provided 
by R. Youngs, written comm., 1996), which is based on empirical data of deep events recorded on 
rock sites. The relations of Crouse (1991) were used because they were for soil sites. It was found 
that the ground motions from Geomatrix (1993) are somewhat smaller than those from Crouse (1991), 
by an amount consistent with soil amplification. These events were placed at a depth of 40 km for 
calculation of ground motions. 
Cascadia subduction zone: For M8.3 events on the subduction zone, two attenuation relations (with 
equal weights) were used following the lead of Geomatrix (1993): 1) Sadigh et al. (1993) for crustal 
thrust earthquakes and 2) Geomatrix (1993) for interface earthquakes. For the M9.0 scenario, Sadigh 
et al. (1993) formulas could not be used since they are invalid over M8.5. Therefore, only Geomatrix 
(1993) was used. Again the values from Geomatrix (1993) were somewhat smaller than the soil 
values in Crouse (1991). 

Development of the USGS Seismic Maps 
347 
ALASKA 
The basic procedure, shown in Figure B9, for constructing the Alaska hazard maps is similar to that 
previously described for the WUS. The maps have been completed and both the maps and 
documentation (USGS, 1998a, 1998b) have been placed on the USGS internet site 
(http://geohazards.cr.usgs.gov/eq/); printing of the maps is in progress. 
Faults. The hazard from nine faults 
was used for the maps (Figure B10). 
Faults were included in the map when 
an estimated slip rate was available. 
The seismic hazard associated with 
faults not explicitly included in the 
map is captured to a large degree by 
the smoothed seismicity model. 
Specific details on the fault 
parameters are given in USGS., 
1997a. All of the faults except one 
were strike-slip faults. 
Recurrence models for faults. As 
was done for the western U.S., faults 
were divided into types A and B. The 
fault treatment was the same as the 
WUS. Type A faults were the Queen 
Charlotte, Fairweather offshore, 
Fairweather onshore, and Transition 
fault. Type B faults included western 
Denali, eastern Denali, Totshunda, 
and Castle Mountain. 
For the type B faults, both 
characteristic and Gutenberg-Richter 
(G-R) models of earthquake 
occurrence were used. These recurrence models were weighted equally. The G-R model accounts for 
the possibility that a fault is segmented and may rupture only part of its length. It was assumed that 
the G-R distribution applies from a minimum moment magnitude of 6.5 up to a moment magnitude 
corresponding to rupture of the entire fault length. 
Special case. The Transition fault was treated as a Type A fault even though its segmentation is 
unknown. Although the rationale for this treatment is documented in USGS, 1998a, it should be 
pointed out that the parameters, such as segmentation and slip rate, associated with this fault are highly 
uncertain. 
Megathrust. The Alaska-Aleutian megathrust was considered in four parts, shown in Figure B11. 
Specific rationale for the use of these boundaries is complex and is described in USGS, 1998a. 
Alaska catalogs. A new earthquake catalog was built by combining Preliminary Determination of 
Epicenter, Decade of North American Geology, and International Seismological Centre catalogs with 
USGS interpretations of catalog reliability. Mueller et al. (1997) describes the processing of the 
catalogs, the removal of aftershocks, and the assignment of magnitudes. 
Figure B9 Seismic hazard models for Alaska. Smoothed 
seismicity models are shown on the left and fault models are 
shown on the right. 
2003 Commentary, Appendix B 
348 
Figure B10 Faults included in the maps. Faults are shown with different line types for 
clarity. Dipping faults are shown as closed polygons. 
Development of the USGS Seismic Maps 
349 
Figure B11 Subduction zones included in the maps 
Attenuation relations for Alaska 
Crustal Events: For spectral response acceleration, two equally-weighted attenuation relations were 
used: (1) Boore, Joyner, and Fumal (BJF; 1997) and (2) Sadigh et al. (1997). For (1) ground motions 
were calculated for a site with average shear-wave velocity of 760 m/sec in the top 30m. For (2) their 
rock values were used. These are recent publication of the attenuations cited for the WUS. The 
attenuations are the same. When calculating ground motions for each fault, the relations appropriate 
for that fault type (e.g, strike slip) were used. All calculations included the variability of ground 
motions. 
Deep events (50 - 80 km): For these deep events, only one attenuation relation was used, the intraslab 
form of Youngs et al. (1997) with a depth fixed at 60 km. 
Deeper events (80 - 120 km): For these deeper events, only one attenuation relation was used, the 
intraslab form of Youngs et al. (1997) with a depth fixed at 90 km. 
Megathrust and Transition Fault: Only one attenuation relation was used, the interslab form of 
Youngs et al. (1997). It should be noted that the use of this attenuation for the Transition fault resulted 
in lower ground motions than would have been obtained using the crustal attenuation equations. 
PROBABILISTIC MAPS 
Two of the probabilistic maps were key to the decisions made by the SDPG for developing the 
maximum considered earthquake ground motion maps. These are the 0.2 sec and 1.0 sec spectral 
response maps for a 2 percent probability of exceedance in 50 years. These are shown in Figures B12 
and B13 respectively. The way in which these maps were used is described in the following sections. 

2003 Commentary, Appendix B 
350 
Figure B12 Probabilistic map of 0.2 sec spectral response acceleration with a 2% 
probability of exceedance in 50 years. The reference site material has a shear wave 
velocity of 750 m/sec. 
Figure B13 Probabilistic map of 1.0 sec spectral response acceleration with a 2% 
probability of exceedance in 50 years. The reference site material has a shear wave 
velocity of 750 m/sec. 

Development of the USGS Seismic Maps 
351 
DEVELOPMENT OF NEHRP MAXIMUM CONSIDERED EARTHQUAKE SPECTRAL 
ACCELERATION MAPS 
The maximum considered earthquake spectral acceleration maps were derived from the 2 percent in 50 
year probabilistic maps shown simplified as Figures B12 and B13 (also see Frankel et al, 1997), 
discussed above, with the application of the SDPG rules also described previously. Additional detail 
in applying the rules is described in this section. The 0.2 sec map is used for illustration purposes. 
The same procedures and similar comments apply for the 1.0 sec map. 
One of the essential features of the SDPG rules was that the recommendations, when applied by 
others, would result in the same maps. This procedures allows the use of engineering judgment to be 
used in developing the maps, as long as those judgments are explicitly stated. This approach will 
simplify modification of the recommendations as knowledge improves. 
It should be noted that although the maps are termed maximum considered earthquake ground motion 
maps. These maps are not for a single earthquake. The maps include probabilistic effects which 
consider all possible earthquakes up to the plateau level. Above the plateau level, the contours are 
included for the deterministic earthquake on each fault (unless the deterministic value is higher than 
the probabilistic values). 
Deterministic contours. The deterministic contours, when included, are computed using the same 
attenuation functions used in the probabilistic analysis. However, the deterministic values are not used 
unless they are less than the probabilistic values. After study of those areas where the plateau was 
reached, the only areas where the deterministic values were less than the probabilistic values were 
located in California and along the subduction zone region of Washington and Oregon. Further study 
indicated that those areas with values in excess of the plateau were located in California. The 
appropriate attenuation for this area were the Boore-Joyner-Fumal attenuation (1993, 1994) and the 
Sadigh et al. (1993) attenuation. 
The form of these attenuations and the distance measures used have an effect on the shape of these 
deterministic contours. Accordingly, they are discussed below. The Boore-Joyner-Fumal equation is: 
logY = bssGss + bRSGRS + b2(M - 6)2 + b4r + b5log(r) + bv(logVs + logVa) 
where: 
Y = ground motion parameter 
M = earthquake magnitude 
bSS, bRS = coefficients for strike-slip and reverse-slip faults, determined by regression and 
different for each ground motion parameter 
GSS = 1.0 for strike-slip fault, otherwise zero 
GRS = 1.0 for reverse-slip fault, otherwise zero 
b2, b3, b4, b5 = coefficients determined by regression, different for each spectral acceleration 
bV = coefficient determined by regression, different for each spectral acceleration 
VA = coefficient determined by regression, different for each spectral acceleration 
VS = shear wave velocity for different site category 
r = (d2 +h2)2 
d = closest horizontal distance from the site of interest to the surface projection of the 
rupture surface, see Figure B14 
h = fictitious depth determined by regression, different for each ground motion 
parameter 
Coefficients determined by regression are tabulated in the reports describing the attenuation equation. 
2003 Commentary, Appendix B 
352 
The Sadigh et al. equation is: 
lnY(T) = F{C1 + C2M + C3(8.5 - M)2.5 + C4ln[D + exp(C5 + C6M)] + Cyln(D + 2)} 
where: 
Y = spectral response acceleration at period T 
M = earthquake magnitude 
C1, C2, C3....C7 = coefficients determined by regression, different for each ground motion parameter 
D = closest distance to the fault rupture surface, see Figure B14 
F = Factor for fault type, 1.0 for strike-slip faults, 1.2 for reverse/thrust faulting, 1.09 
for oblique faults 
The distance measures are shown in Figure B14 and are discussed in more detail below. 
The computation of spectral response (or any ground motion parameter) is a relatively simple matter 
for a specific site (or specific distance from a fault) but can become complex when preparing contours 
since it is difficult to calculate the specific distance at which a particular ground motion occurs. This 
is due to the complexity of the two attenuation functions and the need to combine their results. Since 
the attenuation functions were weighted equally, each contributes equally to the ground motion at a 
site. Deterministic contours were determined by preparing attenuation tables, that is the spectral 
response was computed at various distances from the fault or the fault ends for each earthquake 
magnitude. Contours for specific values were then drawn by selecting the table for the appropriate 
magnitude and determining, using interpolation, the distance from the fault for a given spectral 
acceleration. This procedure required, as a minimum, one attenuation table for each fault. Depending 
on the fault geometry, more than one table was needed. In order to illustrate this the strike-slip fault 
is discussed first, followed by a discussion of dipping faults. 
Development of the USGS Seismic Maps 
353 
Figure B14 Measures of distance for strike-slip and dipping faults. A cross 
section of strike-slip fault is shown in figure (a) and the shape of a typical 
deterministic contour is shown in figure (b). A dipping fault is shown in figure 
(c) and the shape of a typical deterministic contour is shown in figure (d). 
Strike-slip faults: The strike-slip fault, shown in Figure B14a, b is the simplest introduction to 
application of the SDPG rules. The distance measures are shown for each attenuation function in 
Figure B14a. The Boore-Joyner-Fumal equation uses the distance, d4. The term r in equation includes 
d4 and the fictitious depth h. Since h is not zero, r > d4, even if the term y in Figure B14a is zero. The 
Sadigh et al. equation measures the distance, D, as the closest distance to the rupture surface. In this 
case to the top of the rupture. If the depth y is zero, then d4 = D4. 
It makes little difference in the computations if the fault rupture plane begins at the surface or at some 
distance below the surface. For the strike-slip fault the contour for a particular spectral acceleration is 
a constant distant from the fault and the contour is as shown in Figure B14b. One attenuation table 
(including the effects of both attenuation equations) can be used for either side of the fault and at the 
fault ends. 
Dipping faults: The dipping fault, shown in Figures B14c and d, is the most complex case for 
preparing deterministic contours. The distance measures are shown for each attenuation function in 
Figure B14c. As before, it is a simple matter to compute the spectral values at a specific site, but not 
2003 Commentary, Appendix B 
354 
as simple to compute the distance at which a specific spectral acceleration occurs. This is particularly 
true at the end of the fault. 
On the left side of the fault shown in Figure B14c, an attenuation table is prepared, much as in the case 
of the strike-slip fault. This table may also be used to determine the contour around a portion of the 
fault end as shown in Figure B14d. In this case it is simply one-quarter of a circle. 
A separate attenuation table must be prepared for the right side of the fault as shown in Figure B14d. 
Since d or D is measured differently, depending on location x, calculations must keep track of whether 
or not the location is inside or outside of the surface fault projection. Note that the term d is zero when 
the location x falls within the surface projection, but the fictitious depth h is not. Outside the fault 
projection, the distance d is measured from the edge of the projection. The distance D is calculated 
differently, as illustrated in Figure B14c, depending on location but it is always the closest distance to 
the fault rupture surface. 
At the ends of the fault, an attenuation grid was prepared to determine the contour shape shown dotted 
in Figure B14d. The contour in this area was digitized using the gridded values and combined with 
the remainder of the contour determined from the left and right attenuation tables. This need for 
digitizing a portion of the contour greatly increased the time required to prepare each of the contours 
for dipping faults. In short, each dipping fault required 
two attenuation tables and an attenuation grid to 
prepare each deterministic contour. Thus preparation 
of each contour is far more time-consuming than 
preparing a contour for a strike-slip fault. Each 
contour is unsymmetrical around the fault, the amount 
of asymmetry depends on the angle of dip. 
It can be argued that the knowledge of fault locations 
and geometry does not warrant this level of effort. 
However, it was considered necessary in order to 
follow the concept of repeatability in preparing the 
maps. 
Combining deterministic contours: Where two or 
more faults are nearby, as in Figure B15a, the 
deterministic contours were merged (depending on 
amplitudes) as shown in Figure B15b. The merging 
resulted in the sharp corners shown in the figure. 
Although it can be argued that these intersections 
should be smoothed, it was believed that maintaining 
the shape reflected the decision to use deterministic 
contours. 
Combining deterministic and probabilistic 
contours. The SDPG decision to use a combination of 
deterministic and probabilistic contours, although 
simple in principle, led to number of problems in 
preparing the contour maps. 
Figure B16a, b for a single strike-slip fault illustrates 
the concept originally envisioned for combining the deterministic and probabilistic contours. After 
combining the two sets of contours shown in Figure B16a, the maximum considered earthquake 
contours would be as shown in Figure B16b. 
Figure B15 Procedure for combining 
deterministic contours from nearby faults 
Development of the USGS Seismic Maps 
355 
Figure B16 Procedure for obtaining maximum considered earthquake 
ground motion 
In application the situation is more complex, there is frequently more than one fault, with different 
magnitudes, different return times, different fault geometry, and different locations with respect to 
each other. Examples are shown in Figures 17 and 18 which will be discussed later. The effect of the 
variables is illustrated in Figure B16 c and d. The deterministic curve is shown for a single fault with 
a return time much larger than that of the map. The deterministic spectral acceleration is much larger 
than the spectral acceleration resulting from historical seismicity. The probabilistic curve is not 
necessarily symmetrical to the fault. The resulting maximum considered earthquake curve shown in 
Figure B16d is a complex mix of the probabilistic and deterministic curves. There is not always a 
plateau and the curve is not necessarily symmetrical to the fault, even for a strike-slip fault. Simply 
stated, the probabilistic curve considers other sources such as historical seismicity and other faults as 
well as time. The deterministic curve does not consider other sources for this simple example and 
does not consider time. 
The only areas of the United States that have deterministic contours are in California, along the Pacific 
coast through Oregon and Washington, and in Alaska. At first review it can be seen that there are 
several other areas that have contours in excess of the plateau but do not have plateaus. In these areas 
(e.g., New Madrid), the deterministic values exceed the probabilistic ones and thus were not used. 
There were several instances where application of the SDPG rules produced results that appear 
counterintuitive and in other instance produced results that were edited. Two examples from southern 
2003 Commentary, Appendix B 
356 
California are discussed below. Each example is illustrated with a three-part figure. Part (a) shows 
both probabilistic contours (dashed) and deterministic contours (solid) for each fault which is also 
shown. Part (b) shows the maximum considered earthquake results produced by following the SDPG 
rules. Part (c) shows how part (b) was edited for the final map. 
Example 1: The first example in Figure B17 illustrates the occurrence of gaps in the deterministic 
contours around a fault and the halt of a deterministic contour before the end of a fault. When the 
probabilistic contours and deterministic contours shown in Figure B17a are combined, a gap in the 
deterministic contours occurs in the vicinity of 34.6E and 118.8E. Similarly the deterministic contours 
stop prior to the end of the fault around 34.65E and 119.4E. Both of these are shown in Figure B17b. 
Figure B17a Combining contours - Example 1. Both probabilistic and 
deterministic contours are shown. Probabilistic contours are shown dashed. 
Development of the USGS Seismic Maps 
357 
Figure B17b Combining contours - Example 1. Both probabilistic contours are 
merged using strict interpretation of committee rules. 
After study, it is clear that the SDPG rules results in a repeatable, but unusual, set of contours. The 
result does not go along with the concept of accounting for near fault effects with the deterministic 
contours. Because of this undesirable effect, the contours were hand edited to restore the gaps and 
produce the result in Figure B17c. 
All occurrences similar to this were edited to modify the contours so that the deterministic contours 
did not have abrupt breaks or stops before the ends of the fault. 
Figure B17c Combining contours - Example 1. Probabilistic contours are merged 
with deterministic contours using strict interpretation of committee rules with 
subsequent editing. 
2003 Commentary, Appendix B 
358 
Example 2: The second example in Figure B18 illustrates the occurrence of many faults at different 
orientations to each other and with different return times. Merging of the complex set of contours is 
shown in Figure B18b. The contours are greatly simplified. Some small plateaus are shown along the 
150 percent contour, as is a gap along one of the faults around 34.0E and 116.35E. The gap was edited 
as in example 1. The small plateaus were edited out using the judgment that their presence was 
inconsequential (less than a few percent effect on the maps) and unnecessarily complicated an already 
complicated map. 
Another problem created was that some of the faults have portions of the fault, with a specific 
acceleration value, in areas where the contours are less than the fault value. An example occurs with 
the fault labeled 248 in the vicinity of 34.4E and 117.2E. A footnote was added to the maximum 
considered earthquake maps to the effect that the fault value was only to be used in areas where it 
exceeded the surrounding contours. Although other approaches are possible, such as showing the 
unused portion of the fault dashed, the full length of the faults are shown solid in the maps. 
As shown in Figure B18b, a sawtooth contour around 34.15o and 116.3o results from application of 
committee rules. Although this appears to be a candidate for smoothing, it was not done as shown in 
Figure B18c. Once again there are several possible ways to smooth but it was not done in the interest 
of repeatability. 
Figure B18a Combining contours - Example 2. Both probabilistic and 
deterministic contours are shown. Probabilistic contours are shown dotted. 

Development of the USGS Seismic Maps 
359 
Figure B18b Combining contours - Example 2. Probabilistic contours are merged 
using strict interpretation of committee rules. 
Probability level. The maximum considered earthquake spectral acceleration maps use the 2 percent 
in 50 maps as a base; however, the values obtained from the maps are multiplied by 2/3 for use in the 
design equation. This implicitly results in a different probability being used in different areas of the 
United States. The hazard curves shown in Figure B2 are normalized to the 2 percent in 50 year value 
in Figure B19. This figure shows that the slope of the hazard curve varies in different areas of the 
United States. In general, the curves are steeper for CEUS cities than for WUS cities with the WUS 
curves beginning to flatten out earlier than the CEUS cities. Typical curves for a CEUS and WUS city 
are shown in Figure B20. This figure shows than when the 2/3 factor is applied, probabilistic values a 
for WUS location are close to a 10 percent in 50 year value and probabilities for CEUS locations 
reflect a lower probability. 

2003 Commentary, Appendix B 
360 
Figure B18c Combining contours - Example 2. Probabilistic contours are merged 
with deterministic contours using strict interpretation of committee rules with 
subsequent editing. 
Figure B19 Hazard curves for selected cities. The curves are normalized to 2% in 50 years. 

Development of the USGS Seismic Maps 
361 
Figure B20 Effect on the probability level of multiplying the spectral acceleration by 2/3 
Interpolation. Linear interpolation between contours is permitted using the maximum considered 
earthquake maps. To facilitate interpolation, spot values have been provided inside closed contours of 
increasing or decreasing values of the design parameter. Additional spot values have been provided 
where linear interpolation would be difficult. Values have also been provided along faults in the 
deterministic areas to aid in interpolation. 
Hawaii. The Hawaii State Earthquake Advisory Board (HSEAB), in its ballot on the 1997 Provisions, 
proposed different maps from those included in the original BSSC ballot. The HSEABs comments 
were based in part on recent work done to propose changes in seismic zonation for the 1994 and 1997 
Uniform Building Code. The HSEAB also was concerned that in early 1998 the USGS would be 
completing maps that would be more up to date then those included in the original BSSC ballot. 
Essentially, the HSEABs recommendation was that the maps it submitted or the new USGS maps 
should be used for Hawaii. The USGS maps were completed in March 1998 and were reviewed by 
the HSEAB, including proposals for incorporation of deterministic contours where the ground motions 
exceed the plateau levels described previously. The maps were revised in response to review 
comments and the modified design maps are included as part of the Provisions. 
Briefly, the probabilistic maps were prepared using a USGS methodology similar to that used for the 
western United States (Klein et. al.). Two attenuation functions were used: Sadigh as described 
earlier and Munson and Thurber, which incorporates Hawaii data. The Hawaii contour maps 
(Provisions Map 10) are probabilistic except for two areas on the island of Hawaii. The two areas 
(outlined by the heavy border on Map 10) are located on the western and southeastern portion of the 
island. The two areas are defined by horizontal rupture planes at a 9 km depth. Within these zones, 
the spectral accelerations are constant. The western zone uses a magnitude 7.0 event while the 
southwestern zones uses a magnitude 8.2 event. The deterministic values inside the zone and for the 
contours were calculated as described in earlier sections. 
2003 Commentary, Appendix B 
362 
Additional maximum considered earthquake ground motion maps. Maps for Puerto Rico and the 
U.S. Virgin islands were prepared using the USGS methodology described previously with 
modifications and attenuations appropriate for the region as described by Mueller, et. al. The two 
maximum considered earthquake spectral acceleration maps for the region are entirely probabilistic 
since values did not exceed the thresholds requiring incorporation of deterministic values. Although 
new probabilistic maps were not available for Guam and Tutuila, maximum considered earthquake 
maps were required for use by the Provisions. Maximum considered earthquake spectral response 
maps for these areas were prepared as follows. 
Maps for Guam and Tutuila were prepared using the 1994 NEHRP maps. These were for 
approximately 10 percent probability of exceedance in 50 years. The ratio of PGA for 2 percent in 50 
years to 10 percent in 50 years for the new USGS maps is about two. Accordingly maps for these 
areas were converted to 2 percent in 50 year maps by multiplying by two. These maps were then 
converted to spectral maps by using the factors described below. 
A study of the ratios of the 0.2 sec and 1.0 sec spectral responses to PGA was done. Although 
approximate, the ratios were about 2.25 to 2.5 for the 0.2 sec spectral acceleration and about 1.0 for 
the 1.0 sec response. Thus PGA for the above regions was converted to spectral acceleration by 
multiplying PGA by 2.5 for the 0.2 sec response and by 1.0 for the 1.0 sec response. It should be 
noted that the multiplier for the 1.0 sec response varied over a wider range than the 0.2 sec response 
multiplier. It should be used cautiously. 
REFERENCES 
Adams, J., D. H. Weichert, S. Halchuk, and P. W. Basham. 1996. Trial seismic hazard maps of 
Canada-1995: final values for selected Canadian cities, Geological Survey of Canada, Open File 
3283. 
Algermissen, S. T., D. M. Perkins, P. C. Thenhaus, S. L. Hanson, and B. L. Bender. 1990. 
Probabilistic Earthquake Acceleration and Velocity Maps for the United States and Puerto Rico, U.S. 
Geological Survey, Miscellaneous Field Studies Map MF-2120. 
Algermissen, S. T., E. V. Leyendecker, G. A. Bollinger, N. C. Donovan, J. E. Ebel, W. B. Joyner, 
R. W. Luft, and J. P. Singh. 1991. Probabilistic Ground-motion Hazard Maps of Response Spectral 
Ordinates for the United States, in Proceedings of Fourth International Conference on Seismic 
Ordinates for the United States, Stanford Univ., v. II, pp. 687-694. 
Amick , D., and R. Gelinas. 1991. The Search for Evidence of Large Prehistoric Earthquakes along 
the Atlantic Seaboard, Science, v. 251, pp. 655-658. 
Atwater, B .F., and E. Hemphill-Haley. 1996. Preliminary estimates of recurrence intervals for great 
earthquakes of the past 3500 years at northeastern Willapa Bay, Washington, U.S. Geological Survey, 
Open-file report 96-001, 88pp. 
Atwater, B., A. Nelson, J. Clague, G. Carver, D. Yamaguchi, P. Bobrowsky, J. Bourgeois, M. 
Darienzo, W. Grant, S. Palmer, C. Peterson, and M. Reinhart. 1995. Summary of coastal geologic 
evidence for past great earthquakes at the Cascadia subduction zone, Earthquake Spectra, v. 11, pp.1- 
18. 
Black, B. D., W. R. Lund, and B. H. Mayes. 1995. Large earthquakes on the Salt Lake City segment 
of the Wasatch Fault Zone--summary of new information from the South Fork Dry Creek site, Salt 
Lake County, Utah, in Environmental and Engineering Geology of the Wasatch Front Region, W. R. 
Lund, ed., Utah Geological Association Publication 24, 1995. 
Development of the USGS Seismic Maps 
363 
Boore, D. M., W. B. Joyner, and T. E. Fumal. 1993. Estimation of response spectra and peak 
accelerations from western North American earthquakes: an interim report, U.S. Geological Survey 
open-file report 93-509. 
Boore, D. M., W. B. Joyner, and T. E. Fumal. 1994a. Estimation of response spectra and peak 
accelerations from western North American earthquakes: an interim report Part 2, U.S. Geological 
Survey open-file report 94-127. 
Boore, D. M., W. B. Joyner, and T. E. Fumal. 1994b. Ground motion estimates for strike- and 
reverse-slip faults pre-print. 
Boore, D. M., W. B. Joyner, and T. E. Fumal. 1997. Equations for estimating horizontal response 
spectra and peak acceleration from western North American earthquakes: a summary of recent work, 
Seismological Research Letters, v. 68, no. 1, pp. 128-153. 
Chandler, V. W. 1995. The west-central Minnesota earthquake of June 5, 1993: an opportunity to reexamine 
seismicity near the Morris Fault, Seism. Res. Lets, v. 66, no. 6, pp. 113-121. 
Crone, A. J., and K. V. Luza. 1990. Style and Timing of Holocene Surface Faulting on the Meers 
Fault, Southwestern Oklahoma, Geological Society of America Bulletin, v. 102, pp. 1-17. 
Crone, A., M. Machete, L. Bradley and S. Mahal. 1996. Holocene Surface Faulting on the Cheraw 
Fault, Southeastern Colorado, U.S. Geological Survey, MF map, in press. 
Crouse, C. B. 1991. Ground-motion attenuation equations for earthquakes on the Cascadia 
subduction zone, Earthquake Spectra, v. 7, p. 201. 
Doser, D. I. 1985. Source parameters and faulting processes of the 1959 Hebgen Lake, Montana, 
earthquake sequence, J. Gaffes. Res., v. 90, pp. 4537-4556. 
Electric Power Research Institute. 1986. Seismic hazard methodology for the central and eastern 
United States, 10 volumes, EPRI report NP-4726, Electric Power Research Institute, Palo Alto. 
Edge, K. M., B. A. Scheol, and I. Po Lam. 1992. Development of seismic maps for Arizona, Arizona 
Dept. of Transportation, report no. FHA-AZ92-344. 
Frankel, A. 1995. Mapping seismic hazard in the Central and Eastern United States, Seismological 
Research Letters, v. 66, no. 4, pp. 8-21. 
Frankel A., C. Mueller, T. Barnhard, D. Perkins, E. Leyendecker, N. Dickman, S. Hanson, and M. 
Hopper. 1996. National Seismic Hazard Maps, June 1996: Documentation, U.S. Geological Survey, 
Open-file Report 96-532. 
Frankel A., C. Mueller, T. Barnhard, D. Perkins, E. Leyendecker, N. Dickman, S. Hanson, and M. 
Hopper. 1997a. Seismic - Hazard Maps for the Conterminous United States: U.S. Geological 
Survey, Open-file Report 97-130, 12 sheets, scale 1:7,000,000. 
Frankel A., C. Mueller, T. Barnhard, D. Perkins, E. Leyendecker, N. Dickman, S. Hanson, and M. 
Hopper. 1997b. Seismic - Hazard Maps for the California, Nevada and Western Arizona/Utah: U.S. 
Geological Survey, Open-file Report 97-130, 12 sheets, scale 1:2,000,000. 
Geomatrix Consultants. 1993. Seismic margin earthquake for the Trojan site: final unpublished 
report prepared for Portland General Electric Trojan Nuclear Plant, Rainier, Oregon. Note: tables of 
these attenuation relationships are included in Geomatrix , 1995. 
Geomatrix Consultants. 1995. Seismic design mapping state of Oregon, Final Report prepared for 
Oregon Department of Transportation, Salem, Oregon. 
Goldfinger, C., L. D. Kulm, R. S. Yeats, B. Applegate, M. MacKay, and G. R.Cochrane in press. 
Active strike-slip faulting and folding of the Cascadia plate boundary and forearc in central and 
northern Oregon, in Rogers, A. M., W. J. Kockelman, G. Priest, and T. J. Walsh, eds., Assessing and 
2003 Commentary, Appendix B 
364 
reducing earthquake hazards in the Pacific northwest, U.S. Geological Survey Professional Paper 
1560. 
Hyndman, R. D., and K. Wang. 1995. The rupture zone of Cascadia great earthquakes from current 
deformation and the thermal regime, J. Gaffes. Res., v. 100, pp. 22,133-22,154. 
Johnston, A. 1994. Moment magnitude assessment of stable continental earthquakes, Part I: 
instrumental seismicity, pre-print, published with revised relations in Gaffes. J. Int., [1996], v. 124, pp. 
381-414. 
Johnston, A. 1996a. Moment Magnitude Assessment of Stable Continental Earthquakes, Part 2: 
Historical Seismicity, Gaffes. J. Int., v. 125, pp. 639-678. 
Johnston, A. 1996b. Seismic Moment Assessment of Stable Continental Earthquakes, Part 3: 1811- 
1812 New Madrid, 1886 Charleston, and 1755 Lisbon, submitted to Gaffes. J. Int. 
Johnston, A. C., K. J. Coppersmith, L. R. Kanter, and C. A. Cornell. 1994. The earthquakes of stable 
continental regions: assessment of large earthquake potential, EPRI TR-102261, J. F. Schneider, ed., 
Electric Power Research Institute, 309 pp. 
Johnston, A. C., and E. S. Schweig. 1996. The enigma of the New Madrid earthquakes of 1811-1812, 
Annual Review of Earth and Planetary Sciences, v. 24, pp. 339-384. 
Joyner, W. B. 1995. Advocacy Paper #4, Soil is the most appropriate reference site condition, in 
ATC 35-2, Preprints: National Earthquake Ground Motion Mapping Workshop, Applied Technology 
Council. 
Leyendecker, E. V., D. M. Perkins, S. T. Algermissen, P. C. Thenhaus, and S. L. Hanson. 1995. 
USGS Spectral Response Maps and Their Relationship with Seismic Design Forces in Building Codes, 
U.S. Geological Survey, Open-File Report 95-596. 
Lienkaemper, J. J. 1996. Fault parameter compilation for northern California, in preparation. 
Martin, G. R., and R. Dobry. 1994. Earthquake site response and seismic code provisions, NCEER 
Bulletin, v. 8, pp. 1-6. 
Mueller, C., M. Hopper, and A. Frankel. 1996. Preparation of earthquake catalogs for the 1996 
national seismic hazard maps: documentation, U.S. Geological Survey Open-File Report, in review. 
Mueller, C., M. A. Frankel, J. Lahr, and M. Wyss. 1997. Preparation of earthquake catalogs for the 
National Seismic Hazard Maps: Alaska, U.S. Geological Survey Open-File Report, in review. 
Mueller, C., et. al., 2003, Seismic Hazard maps for Puerto Rico: U.S. Geological Survey Map I-abcd. 
Mueller, C., et. al., 2003 Probabilistic Seismic Hazard maps of Puerto Rico: U.S. Geological Survey 
Open-File Report 03-xx,yy p. 
Obermeier, S. F., R. B. Jacobson, J. P. Smoote, R. E. Weems, G. S. Gohn, J. E. Monroe, and D. S. 
Powers. 1990. Earthquake-induced Liquefaction Features in the Coastal Setting of South Carolina 
and in the Fluvial Setting of the New Madrid Seismic Zone, U.S. Geological Survey, Prof. Paper 
1504, 44 pp. 
Obermeier, S. F., P. J. Munson, C. A. Munson, J. R. Martin, A .D. Frankel, T. L. Youd, and E. C. 
Pond. 1992. Liquefaction Evidence for Strong Holocene Earthquakes in the Wabash Valley of 
Indiana-Illinois, Seismological Research Letters, v. 63, pp. 321-336. 
Petersen, M. D., and S. G. Wesnousky. 1994. Fault slip rates and earthquake histories for active 
faults in southern California, Bull. Seism. Soc. Am., v. 84, pp. 1608-1649. 
Petersen, M. D., C. H. Cramer, W. A. Bryant, M. S. Reichle, and T. R. Toppozada. 1996. Preliminary 
hazard assessment for Los Angeles, Ventura, and Orange counties, California, affected by the 17 
January 1994 Northridge earthquake, Bull. Seism. Soc. Am., v. 86, pp. S247-S261. 
Development of the USGS Seismic Maps 
365 
Powell, C., G. Bollinger, M. Chapman, M. Sibol, A. Johnston, and R. Wheeler. 1994. A 
seismotectonic model for the 300-kilometer-long eastern Tennessee seismic zone, Science, v. 264, pp. 
686-688. 
Sadigh, K., C. Y. Chang, N. A. Abrahamson, S. J. Chiou, and M. S. Power. 1993. Specification of 
long-period ground motions: updated attenuation relationships for rock site conditions and adjustment 
factors for near-fault effects, in Proceedings of ATC-17-1 Seminar on Seismic Isolation, Passive 
Energy Dissipation, and Active Control, Applied Technology Council, Redwood City. Note: tables of 
these attenuation relationships are included in Geomatrix, 1995. 
Sadigh, K., C. Y. Chang, J. Egan, F. Makdisi, and R. Youngs. 1997. Attenuation relationships for 
shallow crustal earthquakes based on California strong motion data, Seismological Research Letters, 
v. 68, no. 1, pp. 180-189. 
Satake, K., K. Shimazaki, Y. Tsuji, and K. Ueda. 1996. Time and size of a giant earthquake in 
Cascadia inferred from Japanese tsunami record of January 1700, Nature, v. 379, pp. 246-249. 
Schwartz, D. P., and K. J. Coppersmith. 1984. Fault behavior and characteristic earthquakes: 
examples from the Wasatch and San Andreas fault zones, J. Gaffes. Res., v. 89, pp. 5681-5698. 
Seeber, L., and J. G. Armbruster. 1991. The NCEER-91 Earthquake Catalog: Improved Intensitybased 
Magnitudes and Recurrence Relations for U.S. Earthquakes East of New Madrid, National 
Center for Earthquake Engineering Research, NCEER-91-0021. 
Senior Seismic Hazard Analysis Committee. 1996?. Probabilistic Seismic Hazard Analysis: a 
Consensus Methodology, supported by DOE, NRC, and EPRI. 
Sibol, M. S., G. A. Bollinger, and J. B. Birch. 1987. Estimation of Magnitudes in Central and 
Eastern North America Using Intensity and Felt Area, Bulletin of the Seismological Society of 
America, v. 77, pp. 1635-1654. 
Stover, C. W., and J. L. Coffman. 1993. Seismicity of the United States 1568-1989, revised, U.S. 
Geological Survey Professional Paper 1527. 
Stover, C. W., G. Reagor, and S. T. Algermissen. 1984. United States Earthquake Data File, U.S. 
Geological Survey Open-File Report 84-225. 
Toro, G. R., N. A. Abrahamson, and J. F. Schneider. 1993. Engineering Model of Strong Ground 
Motions from Earthquakes in the Central and Eastern United States, in Guidelines for Determining 
Design Basis Ground Motions, EPRI TR-102293, J. F. Schneider, ed., Electric Power Research 
Institute. 
U.S. Geological Survey National Seismic-Hazard Mapping Project, 1998a, Alaska Seismic-Hazard 
Maps: Documentation, U.S. Geological Survey, Open-file Report, in progress. 
U.S. Geological Survey National Seismic-Hazard Mapping Project, 1998b, Seismic-Hazard Maps of 
Alaska: U.S. Geological Survey, Open-file Report, 6 sheets, scale 1:5,000,000, in progress. 
Ward, S. N. 1994. A multidisciplinary approach to seismic hazard in southern California, Bull. Seism. 
Soc. Am., v. 84, pp. 1293-1309. 
Weichert, D. H. 1980. Estimation of earthquake recurrence parameters for unequal observation 
periods for different magnitudes, Bull. Seism. Soc. Am., 70, pp. 1337-1356. 
Wells, D. L., and K. J. Coppersmith. 1994. New empirical relationships among magnitude, rupture 
length, rupture width, and surface displacement, Bulletin of the Seismological Society of America, v. 
84, pp. 974-1002. 
Wesnousky, S. G. 1986. Earthquakes, Quaternary faults, and seismic hazard in California, J. Gaffes. 
Res., v. 91, pp. 12,587-12,631. 
2003 Commentary, Appendix B 
366 
Wheeler, R. L. 1995. Earthquakes and the Cratonward Limit of Iapetan Faulting in Eastern North 
America, Geology, v. 23, pp. 105-108. 
Working Group on California Earthquake Probabilities. 1995. Seismic hazards in southern 
California: probable earthquakes, 1994 to 2024, Bull. Seism. Soc. Am., v. 85, pp. 379-439. 
Wyman, R.V., M. Karakouzian, V. Bax-Valentine, D. B. Slemmons, L. Peterson, and S. Palmer. 
1993. Geology of Las Vegas, Nevada, Bull. Assoc. of Engineer. Geologists, v. XXX, pp. 33-78. 
Youngs, R., Chiou, W. Silva, and J. Humphrey. 1997. Strong ground motion attenuation relationships 
for subduction zone earthquakes, Seismological Research Letters, v. 68, no. 1, pp. 180-18. 

 367 
The Building Seismic Safety Council (BSSC) 
was established in 1979 under the auspices of 
the National Institute of Building Sciences as 
an entirely new type of instrument for dealing 
with the complex regulatory, technical, social, 
and economic issues involved in developing 
and promulgating building earthquake risk 
mitigation regulatory provisions that are 
national in scope. By bringing together in the 
BSSC all of the needed expertise and all 
relevant public and private interests, it was 
believed that issues related to the seismic 
safety of the built environment could be 
resolved and jurisdictional problems overcome 
through authoritative guidance and assistance 
backed by a broad consensus. 
The BSSC is an independent, voluntary 
membership body representing a wide variety 
of building community interests. Its 
fundamental purpose is to enhance public 
safety by providing a national forum that 
fosters improved seismic safety provisions for 
use by the building community in the planning, 
design, construction, regulation, and utilization 
of buildings. To fulfill its purpose, the BSSC: 
 Promotes the development of seismic 
safety provisions suitable for use 
throughout the United States; 
 Recommends, encourages, and promotes 
the adoption of appropriate seismic safety 
provisions in voluntary standards and 
model codes; 
 Assesses progress in the implementation of 
such provisions by federal, state, and local 
regulatory and construction agencies; 
 Identifies opportunities for improving 
seismic safety regulations and practices and 
encourages public and private organizations 
to effect such improvements; 
 Promotes the development of training and 
educational courses and materials for use by 
design professionals, builders, building 
regulatory officials, elected officials, industry 
representatives, other members of the 
building community, and the public; 
 Advises government bodies on their 
programs of research, development, and 
implementation; and 
 Periodically reviews and evaluates research 
findings, practices, and experience and 
makes recommendations for incorporation 
into seismic design practices. 
The BSSC's area of interest encompasses all 
building types, structures, and related facilities 
and includes explicit consideration and 
assessment of the social, technical, 
administrative, political, legal, and economic 
implications of its deliberations and 
recommendations. The BSSC believes that the 
achievement of its purpose is a concern shared by 
all in the public and private sectors; therefore, its 
activities are structured to provide all interested 
entities (i.e., government bodies at all levels, 
voluntary organizations, business, industry, the 
design profession, the construction industry, the 
research community, and the general public) with 
the opportunity to participate. The BSSC also 
believes that the regional and local differences in 
the nature and magnitude of potentially 
hazardous earthquake events require a flexible 
approach to seismic safety that allows for 
THE COUNCIL: 
PROJECTS AND ACTIVITIES 
 368 
consideration of the relative risk, resources, 
and capabilities of each community. The BSSC 
is committed to continued technical 
improvement of seismic design provisions, 
assessment of advances in engineering 
knowledge and design experience, and 
evaluation of earthquake impacts. It 
recognizes that appropriate earthquake hazard 
risk reduction measures and initiatives should 
be adopted by existing organizations and 
institutions and incorporated, whenever 
possible, into their legislation, regulations, 
practices, rules, codes, relief procedures, and 
loan requirements so that these measures and 
initiatives become an integral part of 
established activities, not additional burdens. 
Thus, the BSSC itself assumes no standardsmaking 
role; rather, it advocates that code- and 
standards-formulation organizations consider 
the BSSCs recommendations for inclusion in 
their documents and standards. 
IMPROVING THE SEISMIC SAFETY OF 
NEW BUILDINGS 
The BSSC program directed toward improving 
the seismic safety of new buildings has been 
conducted with funding from the Federal 
Emergency Management Agency (FEMA). It 
is structured to create and maintain 
authoritative, technically sound, up-to-date 
resource documents that can be used by the 
voluntary standards and model code 
organizations, the building community, the 
research community, and the public as the 
foundation for improved seismic safety design 
provisions. 
The BSSC program began with initiatives 
taken by the National Science Foundation 
(NSF). Under an agreement with the National 
Institute of Standards and Technology (NIST; 
formerly the National Bureau of Standards), 
Tentative Provisions for the Development of 
Seismic Regulations for Buildings (referred to 
here as the Tentative Provisions) was prepared 
by the Applied Technology Council (ATC). 
The ATC document was described as the 
product of a "cooperative effort with the 
design professions, building code interests, and 
the research community" intended to 
"...present, in one comprehensive document, the 
current state of knowledge in the fields of 
engineering seismology and engineering practice 
as it pertains to seismic design and construction 
of buildings." The document, however, included 
many innovations, and the ATC explained that a 
careful assessment was needed. 
Following the issuance of the Tentative 
Provisions in 1978, NIST released a technical 
note calling for " . . . systematic analysis of the 
logic and internal consistency of [the Tentative 
Provisions]" and developed a plan for assessing 
and implementing seismic design provisions for 
buildings. This plan called for a thorough review 
of the Tentative Provisions by all interested 
organizations; the conduct of trial designs to 
establish the technical validity of the new 
provisions and to assess their economic impact; 
the establishment of a mechanism to encourage 
consideration and adoption of the new provisions 
by organizations promulgating national standards 
and model codes; and educational, technical, and 
administrative assistance to facilitate 
implementation and enforcement. 
During this same period, other significant events 
occurred. In October 1977, Congress passed the 
Earthquake Hazards Reduction Act of 1977 (P.L. 
95-124) and, in June 1978, the National 
Earthquake Hazards Reduction Program 
(NEHRP) was created. Further, FEMA was 
established as an independent agency to 
coordinate all emergency management functions 
at the federal level. Thus, the future disposition 
of the Tentative Provisions and the 1978 NIST 
plan shifted to FEMA. The emergence of FEMA 
as the agency responsible for implementation of 
P.L. 95-124 (as amended) and the NEHRP also 
required the creation of a mechanism for 
obtaining broad public and private consensus on 
both recommended improved building design and 
construction regulatory provisions and the means 
to be used in their promulgation. Following a 
series of meetings between representatives of the 
original participants in the NSF-sponsored 
project on seismic design provisions, FEMA, the 
American Society of Civil Engineers and the 
National Institute of Building Sciences (NIBS), 
the concept of the Building Seismic Safety 
Council was born. As the concept began to take 
 369 
form, progressively wider public and private 
participation was sought, culminating in a 
broadly representative organizing meeting in 
the spring of 1979, at which time a charter and 
organizational rules and procedures were 
thoroughly debated and agreed upon. 
The BSSC provided the mechanism or forum 
needed to encourage consideration and 
adoption of the new provisions by the relevant 
organizations. A joint BSSC-NIST committee 
was formed to conduct the needed review of 
the Tentative Provisions, which resulted in 198 
recommendations for changes. Another joint 
BSSC-NIST committee developed both the 
criteria by which the needed trial designs could 
be evaluated and the specific trial design 
program plan. Subsequently, a BSSC-NIST 
Trial Design Overview Committee was created 
to revise the trial design plan to accommodate 
a multiphased effort and to refine the Tentative 
Provisions, to the extent practicable, to reflect 
the recommendations generated during the 
earlier review. 
Trial Designs 
Initially, the BSSC trial design effort was to be 
conducted in two phases and was to include 
trial designs for 100 new buildings in 11 major 
cities, but financial limitations required that the 
program be scaled down. Ultimately, 17 
design firms were retained to prepare trial 
designs for 46 new buildings in 4 cities with 
medium to high seismic risk (10 in Los 
Angeles, 4 in Seattle, 6 in Memphis, 6 in 
Phoenix) and in 5 cities with medium to low 
seismic risk (3 in Charleston, South Carolina, 4 
in Chicago, 3 in Ft. Worth, 7 in New York, and 
3 in St. Louis). Alternative designs for six of 
these buildings also were included. 
The firms participating in the trial design 
program were: ABAM Engineers, Inc.; Alfred 
Benesch and Company; Allen and Hoshall; 
Bruce C. Olsen; Datum/Moore Partnership; 
Ellers, Oakley, Chester, and Rike, Inc.; 
Enwright Associates, Inc.; Johnson and 
Nielsen Associates; Klein and Hoffman, Inc.; 
Magadini-Alagia Associates; Read Jones 
Christoffersen, Inc.; Robertson, Fowler, and 
Associates; S. B. Barnes and Associates; 
Skilling Ward Rogers Barkshire, Inc.; Theiss 
Engineers, Inc.; Weidlinger Associates; and 
Wheeler and Gray. 
For each of the 52 designs, a set of general 
specifications was developed, but the responsible 
design engineering firms were given latitude to 
ensure that building design parameters were 
compatible with local construction practice. The 
designers were not permitted, however, to change 
the basic structural type even if an alternative 
structural type would have cost less than the 
specified type under the early version of the 
Provisions, and this constraint may have 
prevented some designers from selecting the 
most economical system. 
Each building was designed twice  once 
according to the amended Tentative Provisions 
and again according to the prevailing local code 
for the particular location of the design. In this 
context, basic structural designs (complete 
enough to assess the cost of the structural portion 
of the building), partial structural designs (special 
studies to test specific parameters, provisions, or 
objectives), partial nonstructural designs 
(complete enough to assess the cost of the 
nonstructural portion of the building), and 
design/construction cost estimates were 
developed. 
This phase of the BSSC program concluded with 
publication of a draft version of the 
recommended provisions, the NEHRP 
Recommended Provisions for the Development of 
Seismic Regulations for New Buildings, an 
overview of the Provisions refinement and trial 
design efforts, and the design firms' reports. 
The 1985 Edition of the NEHRP 
Recommended Provisions 
The draft version represented an interim set of 
provisions pending their balloting by the BSSC 
member organizations. The first ballot, 
conducted in accordance with the BSSC Charter, 
was organized on a chapter-by-chapter basis. As 
required by BSSC procedures, the ballot 
provided for four responses: "yes," "yes with 
reservations," "no," and "abstain." All "yes with 
reservations" and "no" votes were to be 
 370 
accompanied by an explanation of the reasons 
for the vote and the "no" votes were to be 
accompanied by specific suggestions for 
change if those changes would change the 
negative vote to an affirmative. 
All comments and explanations received with 
"yes with reservations" and "no" votes were 
compiled, and proposals for dealing with them 
were developed for consideration by the 
Technical Overview Committee and, 
subsequently, the BSSC Board of Direction. 
The draft provisions then were revised to 
reflect the changes deemed appropriate by the 
BSSC Board and the revision was submitted to 
the BSSC membership for balloting again. 
As a result of this second ballot, virtually the 
entire provisions document received consensus 
approval, and a special BSSC Council meeting 
was held in November 1985 to resolve as 
many of the remaining issues as possible. The 
1985 Edition of the NEHRP Recommended 
Provisions then was transmitted to FEMA for 
publication in December 1985. 
During the next three years, a number of 
documents were published to support and 
complement the 1985 Provisions. They 
included a guide to application of the 
Provisions in earthquake-resistant building 
design, a nontechnical explanation of the 
Provisions for the lay reader, and a handbook 
for interested members of the building 
community and others explaining the societal 
implications of utilizing improved seismic 
safety provisions and a companion volume of 
selected readings. 
The 1988 Edition 
The need for continuing revision of the 
Provisions had been anticipated since the onset 
of the BSSC program and the effort to update 
the 1985 Edition for reissuance in 1988 began 
in January 1986. During the update effort, 
nine BSSC Technical Committees (TCs) 
studied issues concerning seismic risk maps, 
structural design, foundations, concrete, 
masonry, steel, wood, architectural and 
mechanical and electrical systems, and 
regulatory use. The Technical Committees 
worked under the general direction of a Technical 
Management Committee (TMC), which was 
composed of a representative of each TC as well 
as additional members identified by the BSSC 
Board to provide balance. 
The TCs and TMC worked throughout 1987 to 
develop specific proposals for changes needed in 
the 1985 Provisions. In December 1987, the 
Board reviewed these proposals and decided 
upon a set of 53 for submittal to the BSSC 
membership for ballot. Approximately half of 
the proposals reflected new issues while the other 
half reflected efforts to deal with unresolved 
1985 edition issues. 
The balloting was conducted on a proposal-byproposal 
basis in February-April 1988. Fifty of 
the proposals on the ballot passed and three 
failed. All comments and "yes with reservation" 
and "no" votes received as a result of the ballot 
were compiled for review by the TMC. Many of 
the comments could be addressed by making 
minor editorial adjustments and these were 
approved by the BSSC Board. Other comments 
were found to be unpersuasive or in need of 
further study during the next update cycle (to 
prepare the 1991 Provisions). A number of 
comments persuaded the TMC and Board that a 
substantial alteration of some balloted proposals 
was necessary, and it was decided to submit these 
matters (11 in all) to the BSSC membership for 
reballot during June-July 1988. Nine of the 
eleven reballot proposals passed and two failed. 
On the basis of the ballot and reballot results, the 
1988 Provisions documents were prepared and 
transmitted to FEMA for publication in August 
1988. A report describing the changes made in 
the 1985 edition and issues in need of attention in 
the next update cycle also was prepared, and 
efforts to update the complementary reports 
published to support the 1985 edition were 
initiated. Ultimately, the following publications 
were updated to reflect the 1988 Edition and 
reissued by FEMA: the Guide to Application of 
the Provisions, the handbook discussing societal 
implications (which was extensively revised and 
retitled Seismic Considerations for Communities 
at Risk), and several Seismic Considerations 
handbooks (which are described below). 
 371 
The 1991 Edition 
During the effort to produce the 1991 
Provisions, a Provisions Update Committee 
(PUC) and 11 Technical Subcommittees (TSs) 
addressed seismic hazard maps, structural 
design criteria and analysis, foundations, castin-
place and precast concrete structures, 
masonry structures, steel structures, wood 
structures, mechanical-electrical systems and 
building equipment and architectural elements, 
quality assurance, interface with codes and 
standards, and composite structures. Their 
work resulted in 58 substantive and 45 
editorial proposals for change to the 1988 
Provisions. 
The PUC, under the leadership of Loring 
Wyllie of Degenkolb Enginners, approved 
more than 90 percent of the proposals and, in 
January 1991, the BSSC Board accepted the 
PUC-approved proposals for balloting by the 
BSSC member organizations in April-May 
1991. 
Following the balloting, the PUC considered 
the comments received with "yes with 
reservations" and "no" votes and prepared 21 
reballot proposals for consideration by the 
BSSC member organizations. The reballoting 
was completed in August 1991 with the 
approval by the BSSC member organizations 
of 19 of the reballot proposals. 
On the basis of the ballot and reballot results, 
the 1991 Provisions documents were prepared 
and transmitted to FEMA for publication in 
September 1991. Reports describing the 
changes made in the 1988 Edition and issues in 
need of attention in the next update cycle also 
were developed. 
In August 1992, in response to a request from 
FEMA, the BSSC initiated an effort to 
continue its structured information 
dissemination and instruction/training effort 
aimed at stimulating widespread use of the 
Provisions. The primary objectives of the 
effort were to bring several of the publications 
complementing the Provisions into 
conformance with the 1991 Edition in a 
manner reflecting other related developments 
(e.g., the fact that all three model codes now 
include requirements based on the Provisions) 
and to bring instructional course materials 
currently being used in the BSSC seminar series 
(described below) into conformance with the 
1991 Provisions. 
The 1994 Edition 
The effort to structure the 1994 PUC and its 
technical subcommittees was initiated in late 
1991 chairing the OUC again for this cycle was 
Loring Wyllie. By early 1992, 12 Technical 
Subcommittees were established to address 
seismic hazard mapping, loads and analysis 
criteria, foundations and geotechnical 
considerations, cast-in-place and precast concrete 
structures, masonry structures, steel structures, 
wood structures, mechanical-electrical systems 
and building equipment and architectural 
elements, quality assurance, interface with codes 
and standards, and composite steel and concrete 
structures, and base isolation/energy dissipation. 
The TSs worked throughout 1992 and 1993 and, 
at a December 1994 meeting, the PUC voted to 
forward 52 proposals to the BSSC Board with its 
recommendation that they be submitted to the 
BSSC member organizations for balloting. Three 
proposals not approved by the PUC also were 
forwarded to the Board because 20 percent of the 
PUC members present at the meeting voted to do 
so. Subsequently, an additional proposal to 
address needed terminology changes also was 
developed and forwarded to the Board. 
The Board subsequently accepted the PUCapproved 
proposals; it also accepted one of the 
proposals submitted under the "20 percent" rule 
but revised the proposal to be balloted as four 
separate items. The BSSC member organization 
balloting of the resulting 57 proposals occurred 
in March-May 1994, with 42 of the 54 voting 
member organizations submitting their ballots. 
Fifty-three of the proposals passed, and the ballot 
results and comments were reviewed by the PUC 
in July 1994. Twenty substantive changes that 
would require reballoting were identified. Of the 
four proposals that failed the ballot, three were 
withdrawn by the TS chairmen and one was 
 372 
substantially modified and also was accepted 
for reballoting. The BSSC Board of Direction 
accepted the PUC recommendations except in 
one case where it deemed comments to be 
persuasive and made an additional substantive 
change to be reballoted by the BSSC member 
organizations. 
The second ballot package composed of 22 
changes was considered by the BSSC member 
organizations in September-October 1994. 
The PUC then assessed the second ballot 
results and made its recommendations to the 
BSSC Board in November. One needed 
revision identified later was considered by the 
PUC Executive Committee in December. The 
final copy of the 1994 Edition of the 
Provisions including a summary of the 
differences between the 1991 and 1994 
Editions was delivered to FEMA in March 
1995. 
The 1997 Edition 
In September 1994, NIBS entered into a 
contract with FEMA for initiation of the 39- 
month BSSC 1997 Provisions update effort. 
Late in 1994, the BSSC member organization 
representatives and alternate representatives 
and the BSSC Board of Direction were asked 
to identify individuals to serve on the 1997 
PUC and its TSs. The 1997 PUC, chaired by 
Bill Holmes of Rutherford and Chekene, was 
constituted early in 1995, and 12 PUC 
Technical Subcommittees were established to 
address design criteria and analysis, 
foundations and geotechnical considerations, 
cast-in-place/precast concrete structures, 
masonry structures, steel structures, wood 
structures, mechanical-electrical systems and 
building equipment and architectural elements, 
quality assurance, interface with codes and 
standards, composite steel and concrete 
structures, energy dissipation and base 
isolation, and nonbuilding structures. 
As part of this effort, the BSSC developed for 
the 1997 Provisions a revised seismic design 
procedure. Unlike the design procedure based 
on U.S. Geological Survey (USGS) peak 
acceleration and peak velocity-related 
acceleration ground motion maps developed in 
the 1970s and used in earlier editions of the 
Provisions, the new design procedure involves 
new design maps based on recently revised 
USGS spectral response maps and a process 
specified within the body of the Provisions. This 
task was conducted with the cooperation of the 
USGS (under a Memorandum of Understanding 
signed by the BSSC and USGS) by the Seismic 
Design Procedure Group (SDPG) working with 
the guidance of a five-member Management 
Committee. 
More than 200 individuals participated in the 
1997 update effort, and more than 165 
substantive proposals for change were developed. 
A series of editorial/organizational changes also 
were made. All draft TS, SDPG, and PUC 
proposals for change were finalized in late 
February 1997, and in early March, the PUC 
Chair presented to the BSSC Board of Direction 
the PUCs recommendations concerning 
proposals for change to be submitted to the BSSC 
member organizations for balloting. The Board 
accepted these recommendations, and the first 
round of balloting was conducted in April-June 
1997. 
Of the 158 items on the first ballot, only 8 did not 
pass; however, many comments were submitted 
with no and yes with reservations votes. 
These comments were compiled for distribution 
to the PUC, which met in mid-July to review the 
comments, receive TS responses to the comments 
and recommendations for change, and formulate 
its recommendations concerning what items 
should be submitted to the BSSC member 
organizations for a second ballot. The PUC 
deliberations resulted in the decision to 
recommend to the BSSC Board that 28 items be 
included in the second ballot. The PUC Chair 
subsequently presented the PUCs 
recommendations to the Board, which accepted 
those recommendations. 
The second round of balloting was completed in 
October. All but one proposal passed; however, 
a number of comments on virtually all the 
proposals were submitted with the ballots and 
were immediately compiled for consideration by 
the PUC. The PUC Executive Committee met in 
 373 
December to formulate its recommendations to 
the Board, and the Board subsequently 
accepted those recommendations. 
The PUC concluded its update work by 
identifying issues in need of consideration 
during the next update cycle and technical 
issues in need of study. The final version of 
the 1997 Provisions, including an appendix 
describing the differences between the 1994 
and 1997 edition, was transmitted to FEMA in 
February 1998. The contract for the 1997 
update effort was extended by FEMA to 
September 1999 to permit several 
complementary initiatives to be pursued. 
One of these initiatives resulted in a CD that 
provides all of the design mapping data needed 
for use with the 1997 NEHRP Recommended 
Provisions and International Building Code as 
well as the International Residential Code and 
the NEHRP Guidelines for the Seismic 
Rehabilitation of Buildings. This CD was 
developed for the BSSC by Dr. E. V. 
Leyendecker of the U.S. Geological Survey. It 
permits the user to search either by longitude 
and latitude or by zipcode. Although the CDROM 
is distributed by FEMA and the BSSC, 
he International Code Council was given 
permission to reproduce copies to accompany 
the International Building Code (IBC) and 
International Residential Code (IRC). 
The second initiative resulted in a list of the 
relevant seismic design map data on a countyby-
county basis. One listing identifies 
populated places, state, county, population 
(when available), latitude and longitude, two 
maximum considered earthquake (MCE) 
spectral points (for use with the 1997 NEHRP 
Recommended Provisions, International 
Building Code: two spectral points for the 10 
percent probability in 50 year maps (for use 
with the NEHRP Guidelines for the Seismic 
Rehabilitation of Buildings), and the 
corresponding category for use with the 
International Residential Code. The final 
version of this listing can be sorted 
alphabetically by county and then by place in 
the county. Another listing presents the 
counties for each state and provides the same 
information as in the first listing but uses the 
approximate geographic or centroid 
coordinates to determine the data grid values for 
each county as a whole. These listings are based 
on the USGA developed CD and were assembled 
for the BSSC by Richard McConnell. 
In a somewhat related effort, the BSSC 
commissioned a set of approximately 40 
comparative designs. Each comparative design 
was performed at least three times: once 
according to the proposed 2000 IBC (which is 
being take to represent the 1997 NEHRP 
Recommended Provisions), once according to the 
1991 Provisions (requirements reflected in the 
National Building Code and Standard Building 
Code), and once according to the 1994 Uniform 
Building Code. Performing the study for the 
BSSC were the J. R. Harris and Company and S. 
K. Ghosh Associates, Inc. 
Also developed during this update cycle was the 
BSSC website ----www.bssconline.org . The site 
provides BSSC with a means for posting 
proposals for changes and other information for 
public comment and also is a venue for a host of 
downloadable material including the Provisions 
and Commentary. 
The 2000 Edition 
In September 1997, NIBS entered into a contract 
with FEMA for initiation of the 48-month BSSC 
effort to update the 1997 NEHRP Recommended 
Provisions for Seismic Regulations for New 
Buildings and Other Structures. 
In lieu of the Seismic Design Procedure Group 
(SDPG) used in the 1997 update, the BSSC reestablished 
Technical Subcommittee 1, Seismic 
Design Mapping, used in earlier updates of the 
Provisions. This subcommittee is composed of 
an equal number of representatives from the earth 
science community, including representatives 
from the USGS, and the engineering community. 
An additional 11 subcommittees were formed to 
address seismic design and analysis, foundations 
and geotechnical considerations, cast-in-place 
and precast concrete structures, masonry 
structures, steel structures, wood structures, 
 374 
mechanical-electrical systems and building 
equipment and architectural elements, quality 
assurance, composite steel and concrete 
structures, base isolation and energy 
dissipation, and nonbuilding structures. Two 
ad hoc task groups also were formed: one to 
develop appropriate anchorage requirements 
for concrete/masonry/wood elements and the 
other to develop a simplified procedure for use 
in the lower seismic risk areas. No technical 
subcommittee was established in this update 
cycle to serve specifically as the interface with 
codes and standards; rather, the BSSCs Code 
Resource Support Committee provided for the 
needed liaison between the PUC and the model 
code and standards organizations. 
The first ballot encompassing 146 proposals 
for change to the 1997 Provisions was 
submitted to the BSSC member organizations 
in April, 2000; the ballot deadline in June. The 
proposals for change also were posted for 
comment on the BSSC website. Of the 64 
member organizations who received ballot 
packages, 42 responded. Of the 146 proposals, 
69 passed with no no votes but some yes 
with reservations votes, 71 passed but with 
no and yes with reservations votes, and 6 
did not pass (i.e., received less than 67 percent 
yes and yes with reservations votes). The 
comments submitted with no and yes with 
reservations votes were compiled and 
distributed to the PUC Technical 
Subcommittee chairs. The PUC then met in 
Denver in July 2000 to receive the TSs 
responses to ballot comments and formulate 
recommendations concerning items that need 
to be submitted to the member organizations 
for a second ballot 
In August 2000, PUC Chair William Holmes 
briefed the BSSC Board of Direction on the 
results of the first ballot and recommended that 
17 items be submitted to the membership for a 
second ballot. Ten of the proposals were 
revisions of previous proposals, three were 
new proposals, and four were proposals 
developed by the PUC to clarify concerns 
arising from the first ballot. 
The official second ballot package was mailed 
to BSSC member organizations for voting in 
September- October 2000. Of the 66 BSSC 
member organizations, 42 responded and all 
proposals passed. There were, however, several 
yes with reservations and no votes, and the 
PUC met on October 30-31, 2000, to resolve the 
comments submitted with these votes and to 
formulate recommendations concerning a third 
ballot. 
On November 1, 2000, the PUC chair presented 
the second ballot results to the BSSC Board and 
recommended that several items be submitted to 
the membership for a third ballot. The primary 
purpose of the third ballot was to permit 
integration into the 2000 Provisions of new steel 
requirements resulting from the FEMA-funded 
SAC effort mounted to study damage during the 
Northridge earthquake and of the most current 
version of the American Institute of Steel 
Constructions standard which was expected to 
include many of the SAC requirements. The 
third ballot, which included five proposals, was 
sent to the membership for vote by February 
2001. Of the 65 member organizations, 44 
submitted ballots (67 percent). All five proposals 
passed and the results were reviewed and 
comments resolved by the PUC Executive 
Committee at a meeting in March 2001. 
The PUC chair briefed the BSSC Board on the 
third ballot results on March 6, 2000, and the 
Board unanimously approved the 2000 
Provisions for transmittal to FEMA following a 
final editorial review by the PUC of the 
Provisions document and its accompanying 
Commentary volume. Reports identifying the 
major differences between the 1997 and the 2000 
Editions of the Provisions and describing 
unresolved issues and major technical topics in 
need of further study also were prepared. 
Code-language versions of changes for the 2000 
Provisions for submittal as proposed code 
changes for the 2003 Edition of the IBC were 
developed for the BSSC by S. K. Ghosh 
Associates. 

 375 
The 2003 Edition 
Well before the actual contract between FEMA 
and the BSSC was awarded, planning for the 
2003 Edition was under way. Several major 
where the initial topics of attention. First, in 
January 2001a meeting was held to decide how 
best to handle the diverse subject of 
nonbuilding structures. It was concluded that 
the best solution for the 2003 cycle was to 
recommend that the nonbuilding structures 
technical subcommittee (T S 13) continue but 
have greater representation on the PUC with 
four members. It was also recommended that 
TS 13 form eight subgroups to address major 
nonbuilding structures categories such as 
chimneys, wharves and piers, tanks and 
vessels, etc. 
The second area of concern was a detailed edit 
of the 2000 provisions to eliminate the undue 
repetition and inconsistencies that had crept in 
over the years. This edit performed at the end 
of the 2000 cycle by Michael valley of 
Magnusson and Klemencic. After a thorough 
review of the edited document, this 
Reformatted version was voted on by the 
BSSC membership in October 2001. It was 
accepted and became the basis for the 2003 
update. 
The final issue involved structuring the 2003 
update to reflect the fact that the Provisions 
requirements were being reflected in ASCE 7, 
the IBC, the IRC, and the NFPA 5000. Further 
it seemed likely that the I codes would cover 
most seismic matters by referencing the ASCE 
7. Thus it appeared most reasonable to 
coordinate the BSSC efforts with those of 
ASCE 7 Seismic Task Committee, there by 
relieving the PUC and its TSs of the 
responsibility for maintaining code language. 
Considerable progress has been made on 
integrating ASCE 7 as a full reference standard 
during the 2003 update cycle and it is expected 
that this effort will be completed during the 
next update. 
The proposal for the 2003 update of the 
NEHRP Recommended Provisions was 
submitted to FEMA in June 2001. In order to 
keep the momentum of the update process and 
with the concurrence of the FEMA Project 
Officer, candidates for update committee 
membership were identified and recommended 
membership lists were reviewed and accepted by 
the BSSC Board at a June 2001 meeting along 
with a revised procedures/goals statement for the 
effort developed to reflect the thoughts expressed 
at the BSSC Annual Meeting in March. Letters 
of invitation to serve on the update committees 
were mailed in late June 2001. 
The 2003 Provisions Update Committee (PUC) 
convened for the first time in July in conjunction 
with a meeting of the Joint Correlating 
Committee, which was established for the 2003 
update cycle to eliminate duplication of efforts 
by those working on the Provisions and those 
working on ASCE 7. The PUC Technical 
Subcommittee (TS) chairs identified topics they 
intended to consider during the update and a 
tentative schedule for the project was established. 
FEMA signed the contract with NIBS for the 
2003 update project on September 28, 2001. 
This 30-month contract provides for conduct of a 
base series of tasks and two options. 
A comprehensive edit of the 2000 Provisions 
initiated in early 2000 to eliminate undue 
repetition and inconsistencies and generally make 
the document more user-friendly was completed 
in late summer and reviewed by the PUC. After 
revisions to reflect PUC member comments, the 
draft reformatted document was accepted by the 
BSSC Board for balloting by the BSSC member 
organizations to determine whether the revised 
draft could be used as the base document for the 
remainder of the 2003 update effort. The 
balloting occurred between October and 
December 2001. Forty of the 65 BSSC member 
organizations submitted ballots on the 
reformatted Provisions and the document was 
approved; however, a number of significant 
comments accompanied the ballots. These 
comments were compiled and responses 
formulated. In January 2002, the PUC debated 
resolution of the comments and submitted its 
recommendations to the BSSC Board, which 
accepted the PUC recommendations and 
approved use of the reformatted 2000 Provisions 
 376 
as the base document for the remainder of the 
2003 update effort. 
Work also began in early 2002 on development 
of the new BSSC website that is expected to 
permit Technical Subcommittee and PUC 
members to develop, review, and vote on 
proposals in an interactive electronic 
environment and that also will permit the 
BSSC member organizations to receive 
proposals and submit their ballots 
electronically. 
Proposals for change to be submitted to the 
PUC for ballot were submitted in late August 
2002 and were mailed to the PUC for balloting 
on September 11, 2002. Completed ballots 
were due in mid-October, and the results were 
compiled for review/response by the relevant 
Technical Subcommittee in preparation for 
review by the full PUC. The PUC then met in 
Washington, D.C., on November 7-8 and 
formulated its recommendations for the BSSC 
Board concerning proposals to be submitted to 
the BSSC member organizations for ballot. Of 
the 77 proposals initially submitted by its 
technical subcommittees, the PUC 
recommended to the Board that 54 proposals 
be submitted to the BSSC member 
organizations for ballot but that this balloting 
not occur until all proposals for change for the 
2003 Provisions are completed. The Board 
accepted this recommendation, and remaining 
proposals were scheduled to be submitted for 
PUC review by April 1, 2003. 
Approximately 90 proposals were submitted 
for a mail ballot by the PUC. This balloting 
was completed in early June and the PUC met 
on June 15-17, 2003, to resolve comments and 
formulate its recommendations concerning 
which of this second batch of proposals should 
be submitted to the membership. The BSSC 
Board received and accepted the PUC 
recommendations on June 18. 
Ninety-nine new proposals (those submitted to 
the PUC by mail plus a number of PUC 
proposals developed at the meeting) were 
reviewed and voted on by the PUC at a threeday 
meeting held in San Diego, California, in 
June15-17, 2003. Of these, 84 were accepted by 
the PUC, many with revisions, and subsequently 
submitted to the BSSC Board with the 
recommendation that they be added to the 54 
proposals approved earlier and submitted to the 
BSSC member organizations for ballot. The 
Board accepted the PUC recommendation and 
the ballot package (composed of the ballot sheet, 
proposals, composites of the reformatted 
Provisions and Commentary, and the comments 
and responses on each proposal) was sent to the 
representatives and alternates of the 63 BSSC 
member organizations on August 1, 2003. 
Ballots were due October 1. 
The member organization votes were tallied and 
comments were forwarded to the appropriate 
PUC technical subcommittee chairs in mid- 
October in preparation for a November meeting 
of the PUC at which ballot comments were 
addressed. Given that the contract with FEMA 
requires delivery of the consensus approved 2003 
Provisions and Commentary in March 2004, 
another ballot will not be possible; therefore, the 
BSSC Board authorized the PUC to resolve, if 
possible, comments on proposals that have 
passed the membership ballot and to consider any 
proposals for which comments cannot be 
resolved as items for reconsideration in the next 
update cycle. 
The PUC met on November 20-21, 2003, to 
review the proposals for change. Approximately 
130 proposals received the required two-thirds 
affirmative votes; with approximately half of 
those requiring some revisions in response to 
comments. On November 22, the PUC chair 
presented the results to the BSSC Board. The 
Board addressed two contentious issues at the 
request of the PUC and accepted the PUC 
recommendations regarding the changes to be 
made for the 2003 edition of the Provisions and 
Commentary. The final draft is now being 
assembled for review by the PUC and it will be 
officially delivered to FEMA by the end of April 
2004. 
Planning for the next update cycle is beginning 
and a small task group met on November 19, 
2003, to discuss how to structure the next 
Provisions update cycle to adopt ASCE 7 by 
 377 
reference. It also appears that the PUC will be 
somewhat smaller in the next cycle and there 
will be fewer technical subcommittees. Ad 
hoc issue committees will be appointed on an 
as-needed basis to address research needs and 
develop emerging technologies. Coupled with 
this streamlining, it is anticipated that the next 
edition of the Provisions will be issued in 
2008, rather than 2006, to better mesh with the 
codes and standards development schedules. 
CODE RESOURCE DEVELOPMENT 
AND SUPPORT 
In mid-1996, FEMA asked the BSSC to 
initiate an effort to generate a code resource 
document based on the 1997 Provisions for use 
by the International Code Council (ICC) in 
adopting seismic provisions for the first edition 
of the International Building Code (IBC) to be 
published in 2000. The Code Resource 
Development Committee (CRDC) appointed to 
conduct this effort met several times over the 
next year and the CRDC-developed draft 
requirements were presented to the ICCs IBC 
Structural Subcommittee in March 1997. 
Subsequently, the CRDC met to develop 
comments on the IBC working draft to be 
submitted to the ICC in preparation for an 
August 1997 public comment forum. These 
comments generally reflected actions taken by 
the PUC in response to comments submitted 
with the first ballot on the changes proposed 
for the 1997 Provisions as well as CRDC 
recommendations concerning changes made by 
the IBC Structural Subcommittee in the 
original CRDC submittal. CRDC 
representatives attended the August forum to 
support the CRDC recommendations. 
After issuance of the first draft of the IBC in 
November 1997, the CRDC met to prepare 
code change proposals that reflected the 
final version of the 1997 Provisions for 
submittal in January 1998. The CRDC then 
met for the last time as a committee in March 
1998 to review the compilation of IBC code 
change proposals issued by the ICC and to 
develop a strategy for supporting the code 
change proposals it had developed at an IBC 
public hearing in April. In addition, the IBC 
Structural Subcommittee asked for CRDC input 
concerning all the seismic-related code change 
proposals and these comments were summarized 
and transmitted to the IBC group for its 
consideration. 
An eight-member Code Resource Support 
Committee (CRSC) then was established to 
support the Provisions-based requirements 
through the remainder of the adoption process 
and to provide for needed liaison with the 2000 
Provisions development work. A CRSC 
Technical Advisory Group (TAG) composed of 
representatives of the 2000 PUC and the various 
materials interests also was established to support 
the CRSC. The first task of the CRSC was to 
deal with one major issue that arose at the April 
hearing at which several code change proposals 
concerning the draft IBC (and 1997 Provisions 
based) response modification factors and limits 
of applicability of certain structural systems were 
discussed. At the suggestion of a CRDC 
representative at the hearing, the proponents of 
those code changes agreed to withdraw their 
proposals to permit discussion of their technical 
merit outside the forum of the public hearing 
process. To this end, the CRSC invited these 
code change proponents as well as 
representatives of the various construction 
industry materials associations to an August 1998 
meeting at which the group formulated a 
consensus opinion on an appropriate series of 
code change proposals that could be submitted to 
replace those withdrawn in April. Additional 
topics also were discussed and a total of 13 codechange 
proposals were drafted. 
In September 1998, the 2000 PUC Executive 
Committee was briefed on these code-change 
proposals, most of which were accepted by the 
PUC as items to be considered during the 2000 
update effort; however, five items were deemed 
to be significant departures from the 1997 
Provisions and required a vote by the full PUC. 
This balloting concluded in early October with 
all items achieving consensus approval. The 
CRSC then finalized all 13 of its code change 
proposals and submitted them to the ICC in late 
October 1998. 
 378 
In January and February 1999, the CRSC met 
with its Technical Group to consider the 
proposed changes to the International Building 
Code seismic provisions that would be debated 
at March 1999 hearings. The CRSC chair and 
several member participated in the hearings on 
behalf of the CRSC. 
An International Residential Code Task Group 
established within the CRDC in late-1997 has 
provided the ICC committee developing the 
International Residential Dwelling Code (IRC) 
with input concerning seismic requirements 
reflecting the 1997 Provisions, and these 
requirements generally were reflected in the 
draft IRC. The activities of this task group 
have paralleled those of the CRDC/CRSC with 
the IBC and representatives attended the IRC 
July 1998 public hearing in Kansas City. At 
this hearing, agreement was reached on the 
seismic map to be included in the IRC; this 
map subsequently was prepared for the BSSC 
by USGS and submitted to the ICC for 
inclusion in the final draft of the IRC. The task 
group met in February 1999 to review 
proposed code changes and prepare for the 
March ICC hearings. 
The CRSC chair and several CRSC members 
represented the group at the joint annual 
conference of BOCA, ICBO, and SBCCI held 
in September 1999 in St. Louis. Overall, the 
CRSC was successful in that almost all 
challenges to the seismic provisions were 
decided in favor of the CRSC position and the 
seismic provisions in both the 2000 
International Building Code and the 
International Residential Code reflect the 1997 
NEHRP Recommended Provisions. 
In preparation for the ICC hearings to be held 
in Birmingham, Alabama, in April 2000, the 
CRSC and its Technical Group reviewed the 
code changes and met via telephone 
conference calls in March 2000 to discuss the 
proposals. The CRSC chair and several other 
CRSC members attended the hearings. With 
respect to the International Building Code, the 
CRSC had specific positions on 41 proposals. 
Of these proposals, 35 were decided in the 
direction CRSC favored and two that the CRSC 
opposed were withdrawn. During the hearings 
on the International Residential Code, the CRSC 
had specific positions on 12 proposals. Eight of 
these proposals were decided in favor of CRSC's 
position and one was withdrawn at CRSC's 
request. 
In late September 2000, NIBS entered into a 
contract with FEMA to fund further code support 
work by the BSSC. Thus, the 2001 CRSC was 
reconstituted to include additional members and 
two special task groups; one to focus on the IRC, 
and one to focus on the NFPA code. The 
expanded CRSC and its Technical Advisory 
Group (TAG) reviewed the proposals for change 
to the IBC and IRC in preparation for the hearing 
held in Portland, Oregon, in late March 2001. 
During a February 23 conference call, the CRSC 
formulated its position on the proposed changes 
to the IRC. At a meeting on March 5, with its 
TAG, the CRSC decided upon its positions on 
the proposed changes to the IBC. The CRSC 
chair and several CRSC members attended the 
hearing. 
The CRSCs NFPA Task Group members 
attended meetings of the NFPA Technical 
Correlating Committee (TCC) and Structures and 
Construction Committee. In addition, the CRSC 
representative to the TCC has been appointed by 
that committee as its representative to the 
Performance Task Group to the Fundamentals 
Committee. 
The 2001 CRSC met in Denver in July 2001 to 
review draft code change proposals based on the 
changes made for the 2000 Provisions. As noted 
above, S. K. Ghosh Associates, Inc., prepared 
drafts of the IBC-related proposals and Kelly 
Cobeen and Alan Robinson developed the IRC 
proposals. These proposals were then revised in 
response to CRSC comments and, as directed by 
the BSSC Board, sent to the BSSC member 
organizations for comment. 
The CRSC met in October 2001 to review the 
comments received and to address other coderelated 
matters including the need for additional 
changes identified during work on ASCE 7 and 
CRSC work on NFPA 5000. CRSC-approved 
 379 
code changes then were officially submitted to 
the ICC in November. 
Several CRSC members presented an 
educational workshop on the enforcement 
implications for code officials of adoption of 
the IBC/IRC at the BOCA Annual Business 
Meeting in September 2001. 
The CRSC initiated its review of those 
proposals for change to the IBC and IRC 
affecting seismic matters in mid-February 
2002 in preparation for a mid-March meeting 
at which the group formulated its official 
position on these proposals and identified 
which CRSC members would represent the 
committee at the ICC hearings scheduled for 
April 2002. Several members of the CRSC 
attended the hearings and, overall, the groups 
positions on specific changes tended to prevail. 
In mid-September 2002, the CRSC convened 
via telephone to review the final action agenda 
for the ICC Codes Forum to be held in Ft. 
Worth, Texas, the first week in October. Plans 
were also made for CRSC representation at the 
meeting. 
In early March, BSSC was informed that a 
proposed educational workshop on the 
enforcement implications for code officials of 
adoption of the IBC/IRC had been accepted for 
presentation at the Ft. Worth meeting and in 
mid-September the individuals involved in the 
presentation convened via telephone to refine 
plans for the presentation. 
Several CRSC members attended the ICC 
Codes Forum in Fort Worth in October 2002. 
The educational session on IBC/IRC code 
enforcement implications also was conducted 
twice and was well received. During late 
2002, CRSC representatives attended code 
adoption meetings in Kentucky and South 
Carolina. 
The CRSC met in February 2003 to review 
already-accepted proposals for change for the 
2003 Provisions to determine whether any 
should be submitted as proposals for change 
for the International Building Code or 
International Residential Code. The group 
decided that it would submit only one proposal 
for change  i.e., one that would change the map 
for the International Residential Code in the 
central states and the southeast to reduce the area 
in which substantial seismic requirements would 
prevail. 
The CRSC also has nominated several 
individuals to represent CRSC/FEMA interests 
on several technical committees involved in the 
National Fire Protection Association NFPA 5000 
code change process. At FEMAs request, the 
CRSC also nominated an individual to represent 
CRSC/FEMA interests on the NFPA committee 
responsible for two manufactured housing 
standards. 
The CRSC met in June 2003 in conjunction with 
the BSSC Annual Meeting and formulated plans 
for review of proposals in preparation for the ICC 
hearings in September and for an effort to 
develop a change proposal for submittal to the 
ASCE 7 Seismic Task Committee that will 
present a reformatted version of the ASCE 7 
seismic requirements intended to be more user 
friendly and to reflect the reformatting work done 
for the 2003 edition of the NEHRP 
Recommended Provisions. 
Following individual review and input 
concerning the proposals for change to the IBC 
and IRC affecting the seismic requirements, the 
CRSC convened via telephone in August to 
determine the positions to be taken by the CRSC 
representatives who would attend the hearings. 
Subsequently, six CRSC members represented 
the group at the IBC portion of the hearings and 
three, at the IRC portion. As has consistently 
been the case, the positions taken by the CRSC 
tend to prevail with the relevant ICC committees 
overseeing the hearings. 
In July 2003, a representative of the CRSC 
participated in a hearing on IBC adoption in the 
state of Tennessee. In addition, cost information 
developed earlier to show the impact of the 2000 
IBC seismic requirements on costs of typical 
buildings over the costs for the same structures 
constructed under prevailing codes in a number 
of geographic areas was provided to individuals 
 380 
in Tennessee. Planning is under way for the 
development of additional designs that will 
compare the cost impact of the IBC seismic 
requirements over the prevailing code 
requirements for typical buildings in the 
central and southern states. 
Development of the code change proposal for 
ASCE 7 was completed in January and 
officially submitted to the ASCE 7 Seismic 
Task Committee. In addition, the CRSC 
convened in March 2004 to review the public 
comments on ICC code change proposals. 
Few of the comments focused on seismic 
issues; consequently, only two CRSC 
representatives will attend the ICC meeting in 
May. CRSC representatives continue to serve 
on a number of NFPA 5000 technical 
committees and CRSC participants have been 
helpful in drafting a proposal on anchorage of 
manufactured housing to resist earthquake 
ground motions.. 
FEMA also has entered into a new contract 
with NIBS to support the BSSCs codes and 
standards work through FY 2004 and, through 
options, through FY 2006. FEMA currently is 
considering the BSSC proposal for funding for 
the BSSCs code development and support 
functions through FY 2004. 
INFORMATION DISSEMINATION 
The BSSC continues in its efforts to stimulate 
widespread use of the Provisions. In addition 
to the issuance of a variety of publications that 
complement the Provisions, over the past 
decade the BSSC has developed materials for 
use in and promoted the conduct of a series of 
seminars on application of the Provisions 
among relevant professional associations. 
In September 1997, NIBS entered into a 60- 
month indefinite quantity contract with FEMA 
for conduct of the BSSCs information 
dissemination. The first task orders issued 
under the contract charge the BSSC to increase 
its capability to respond to requests for 
technical assistance relating to the Provisions, 
to increase its capability to provide more general 
technical assistance and information in a 
coordinated and proactive manner and using all 
communication media including its website, to 
review existing complementary publications and 
educational seminar materials not already revised 
in whole or in part to reflect the 1997 Provisions 
and to prepare a plan to bring them into 
conformance with the substantive content of the 
1997 Provisions if such is deemed appropriate or 
to develop different documents aimed at 
changing audiences, to revise the course 
materials including the Guide to Application of 
the Provisions, an instructors manual and slide 
set, and a student manual to reflect the 1997 
Provisions and the code requirements based on 
the Provisions, to prepare and implement a plan 
to market the instructional materials and 
subsequently conduct an ongoing series of 
instructional (both technical and nontechnical) 
training seminars on an as-requested basis, to 
continue to promote and encourage the use of the 
Provisions by the nation's model code 
organizations and their adoption by local 
jurisdictions, and to continue to conduct activities 
to increase the general awareness of the 
earthquake risks in different regions throughout 
the country and the need to use local building 
codes that are substantially equivalent with the 
Provisions. 
Because of unanticipated delays in preparation of 
the new Guide and instructional materials, the 
decision was made in early 2001 that the work 
should focus instead on the 2000 edition of the 
Provisions. Since that time, the Guide document 
has been developed, finalized, and reviewed by 
the BSSC subcontractor conducting the project 
and the instructional materials have been pilot 
tested in several venues including the 2001 and 
2002 Multihazard Building Summer Design 
Institutes held in July 2001 and 2002 at the 
Emergency Management Institute. 
The final draft of the new Guide to Application of 
the 2000 NEHRP Recommended Provisions was 
received in autumn 2002 and was sent to the 
Provisions Update Committee, the group 
possessing the greatest in-depth understanding of 
the Provisions, and selected BSSC Board 
members for review. This review was completed 
 381 
in March 2003 and changes were made in 
response to reviewer comments. The final 
draft was then submitted by the BSSC 
subcontractor in April and the material was 
pilot tested in various venues (including the 
Emergency Management Institutes 
Multihazard Building Design Summer Institute 
in July-August 2003). 
The final copy edit of the Guide was 
completed in October 2003 and an effort was 
mounted to extend the life of the document, 
originally prepared to reflect the 2000 edition 
of the Provisions, by integrating cross 
references to the relevant section numbers in 
the 2003 Provisions and by integrating notes 
that focus on changes that will be made for the 
2003 Provisions (mention of those proposals 
that are not approved will be removed before 
FEMA printing of the document). With the 
final decisions now made about the 2003 
Provisions, this effort will be completed by the 
end of April 2004. 
Funded by FEMA in September 2002 was an 
18-month effort to develop an up-to-date 
version of an earlier FEMA publication, Home 
Builders Guide to Seismic-Resistant 
Construction, and to update an earlier BSSC 
publication, Nontechnical Explanation of the 
NEHRP Recommended Provisions, to reflect 
the 2000/2003 Provisions. Work on both these 
documents is well under way and a briefing on 
the plans for the Home Builders was presented 
at the BSSC Annual Meeting in June. These 
two documents in combination with the Guide 
to Application and associated educational 
materials provide the resources needed to 
familiarize a large segment of the building 
community with the Provisions. 
In September 2003, FEMA issued an 
additional task order to fund BSSC 
information dissemination efforts through FY 
2004. 
IMPROVING THE SEISMIC SAFETY OF 
EXISTING BUILDINGS 
Guidelines/Commentary Development 
Project 
The 1997 NEHRP Guidelines for the Seismic 
Rehabilitation and Commentary volumes and 
1997 map packet (which also include maps 
referenced in the NEHRP Recommended 
Provisions for New Buildings and Other 
Structures) are readily available as are two 
companion volumes  Planning for Seismic 
Rehabilitation: Societal Issues (FEMA 275) and 
Example Applications of the NEHRP Guidelines 
for the Seismic Rehabilitation of Buildings 
(FEMA 276). 
Case Studies Project 
The case studies project was an extension of the 
multi-year project leading to publication of the 
NEHRP Guidelines for the Seismic 
Rehabilitation of Buildings and its Commentary 
in late 1997. The project is expected to 
contribute to the credibility of the Guidelines by 
providing potential users with representative realworld 
application data and to provide FEMA 
with the information needed to determine 
whether and when to update the Guidelines. The 
final report on the project was delivered to 
FEMA in September 1999 and is now available 
as FEMA 343, Case Studies: An Assessment of 
the NEHRP Guidelines for the Seismic 
Rehabilitation of Buildings. 
Guidelines Training Seminars 
In August 1997, NIBS entered into a contract 
with FEMA for the design and conduct of a 
series of technical training seminars to transfer 
the technology and information contained in the 
Guidelines to structural and architectural 
engineers (whether in 
private or government practice, representing 
organizations both large and small); to local 
building officials and technical staffs, interested 
contractors, and mitigation officials, where 
applicable; and to engineering educators and 
students in institutions offering seismic design 
curricula. Conceptually, the seminar curriculum 
will take the form of a series of modules that will 
permit it to be adapted for use with a variety of 
audiences. 
The Applied Technology Council, under contract 
to the BSSC, developed the seminar program 
 382 
syllabus and other instructional materials. To 
date, approximately 2000 structural engineers 
have attended seminars on the NEHRP 
Guidelines for \the Seismic Rehabilitation of 
Buildings. Being conducted for FEMA by the 
BSSC with the assistance of the Applied 
Technology Council, two-day seminars have 
been held in San Diego; Salt Lake City; 
Portland, Oregon; Los Angeles; Seattle; New 
York City; Oakland; St. Louis; Charleston, 
South Carolina; Chicago, Illinois; Sacramento, 
California; and Washington, D.C. 

 383 
BSSC MEMBER ORGANIZATIONS 
(* indicates affiliate nonvoting member) 
AFL-CIO Building and Construction Trades 
Department. 
American Concrete Institute 
American Consulting Engineers Council 
American Forest and Paper Association 
American Institute of Architects 
American Institute of Steel Construction 
American Iron and Steel Institute 
American Society of Civil Engineers 
American Society of Civil Engineers--Kansas 
City Chapter 
American Society of Mechanical Engineers 
American Welding Society 
APA - The Engineered Wood Association 
Applied Technology Council 
ASHRAE ,Inc. 
Associated General Contractors of America 
Association of Engineering Geologists 
Association of Major City Building Officials 
Bay Area Structural, Inc.* 
Brick Industry Association 
Building Owners and Managers Association 
International 
Building Technology, Incorporated* 
California Geotechnical Engineers Association 
California Seismic Safety Commission 
Canadian National Committee on Earthquake 
Engineering 
City of Hayward, California* 
Concrete Masonry Association of California 
and 
Nevada 
Concrete Reinforcing Steel Institute 
Concrete Reinforcing Steel Institute 
Division of state Architect (California) 
Earthquake Engineering Research Institute 
Felten Engineering Group, Inc.* 
General Services Administration Seismic 
Program 
Hawaii State Earthquake Advisory Board 
H&H Group, Inc.* 
HLM Design* 
Institute for Business and Home Safety 
Interagency Committee on Seismic Safety in 
Construction 
International Code Council 
International Masonry Institute 
LaPay Consulting, Inc.* 
Masonry Institute of America 
Metal Building Manufacturers Association 
Mid-America Earthquake Center 
National Association of Home Builders 
National Concrete Masonry Association 
National Conference of States on Building 
Codes 
and Standards 
National Council of Structural Engineers 
Associations 
National Elevator Industry, Inc. 
National Fire Sprinkler Association 
National Institute of Building Sciences 
National Ready Mixed Concrete Association 
Portland Cement Association 
Precast/Prestressed Concrete Institute 
Rack Manufacturers Institute 
Santa Clara University 
Square D Company* 
Steel Deck Institute, Inc. 
Steel Joist Institute* 
Structural Engineers Association of California 
Structural Engineers Association of Central 
California 
Structural Engineers Association of Colorado 
Structural Engineers Association of Illinois 
Structural Engineers Association of Kentucky 
Structural Engineers Association of Northern 
California 
Structural Engineers Association of Oregon 
Structural Engineers Association of San Diego 
Structural Engineers Association of Southern 
California 
Structural Engineers Association of Texas 
Structural Engineers Association of Utah 
Structural Engineers Association of 
Washington 
The Masonry Society 
U.S. Army CERL 
Vibration Mountings and Controls* 
Western States Clay Products Association 
Western States Structural Engineers 
Association 
Wire Reinforcement Institute, Inc. 
 384 
BUILDING SEISMIC SAFETY COUNCIL 
PUBLICATIONS 
Available free from the Federal Emergency Management Agency at 1-800-480-2520 (order by 
FEMA Publication Number). For detailed information about the BSSC and its projects, 
contact: BSSC, 1090 Vermont Avenue, N.W., Suite 700, Washington, D.C. 20005 Phone 202- 
289-7800; Fax 202-289-1092; e-mail ctanner@nibs.org 
NEW BUILDINGS PUBLICATIONS 
The NEHRP (National Earthquake Hazards Reduction Program) Recommended Provisions for 
Seismic Regulations for New Buildings, 2003 Edition, 2 volumes and maps, FEMA 450 (issued as a 
CD with only limited paper copies available). 
The NEHRP (National Earthquake Hazards Reduction Program) Recommended Provisions for 
Seismic Regulations for New Buildings, 2000 Edition, 2 volumes and maps, FEMA 368 and 369 
Guide to Application of the 1991 Edition of the NEHRP Recommended Provisions in Earthquake 
Resistant Building Design, Revised Edition, 1995,  new edition to be issued as FEMA 451 in 
preparation 
A Nontechnical Explanation of the NEHRP Recommended Provisions, Revised Edition, 1995, FEMA 
99  new edition in preparation. 
Seismic Considerations for Communities at Risk, Revised Edition, 1995, FEMA 83  new edition 
expected to be published in late 1999 or early 2000 
Seismic Considerations: Apartment Buildings, Revised Edition, 1996, FEMA 152 
Seismic Considerations: Elementary and Secondary Schools, Revised Edition, 1990, FEMA 149 
Seismic Considerations: Health Care Facilities, Revised Edition, 1990, FEMA 150 
Seismic Considerations: Hotels and Motels, Revised Edition, 1990, FEMA 151 
Seismic Considerations: Office Buildings, Revised Edition, 1996, FEMA 153 
Societal Implications: Selected Readings, 1985, FEMA 84 
EXISTING BUILDINGS 
NEHRP Guidelines for the Seismic Rehabilitation of Buildings, 1997, FEMA 273 
NEHRP Guidelines for the Seismic Rehabilitation of Buildings: Commentary, 1997, FEMA 274 
Case Studies: An Assessment of the NEHRP Guidelines for the Seismic Rehabilitation of Buildings, 
1999, FEMA 343 
Planning for Seismic Rehabilitation: Societal Issues, 1998, FEMA 275 
Example Applications of the NEHRP Guidelines for the Seismic Rehabilitation of Buildings, 1999, 
FEMA 276 
NEHRP Handbook of Techniques for the Seismic Rehabilitation of Existing Buildings, 1992, FEMA 
172 
 385 
NEHRP Handbook for the Seismic Evaluation of Existing Buildings, 1992, FEMA 178 
An Action Plan for Reducing Earthquake Hazards of Existing Buildings, 1985, FEMA 90 
MULTIHAZARD 
An Integrated Approach to Natural Hazard Risk Mitigation, 1995, FEMA 261/2-95 
LIFELINES 
Abatement of Seismic Hazards to Lifelines: An Action Plan, 1987, FEMA 142 
Abatement of Seismic Hazards to Lifelines: Proceedings of a Workshop on Development of An Action 
Plan, 6 volumes: 
Papers on Water and Sewer Lifelines, 1987, FEMA 135 
Papers on Transportation Lifelines, 1987, FEMA 136 
Papers on Communication Lifelines, 1987, FEMA 137 
Papers on Power Lifelines, 1987, FEMA 138 
Papers on Gas and Liquid Fuel Lifelines, 1987, FEMA 139 
Papers on Political, Economic, Social, Legal, and Regulatory Issues and General Workshop 
Presentations, 1987, FEMA 143 
(February 2004)